316691663 pile-foundations-lecture-note-1-pdf

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6.0 Construction and selection of deep foundations
Types of Foundation
The foundations can be divided into two main types, namely: Shallow foundations; and Deep
foundations. Terzaghi categorized foundations into the above categories based on the depth of the
foundation below the existing ground surface and classified the deep foundations as the foundations,
whose depth is more than the width of the foundation. However, classification of the foundations based
on the Terzaghi’s concept serves very little purpose in design and construction of the foundations. For
example, an individual footing foundation, having a depth of embedment more than the width, is
designed and constructed in the same way as a footing, whose depth of embedment is less than the
width. Therefore, for engineering purposes foundations should be classified so that there is a clear
difference between the design and construction of the two types. For this purpose, classification of the
foundations based on the load transfer mechanism to the soil or rock is more appropriate. According to
this classification, as shown in Figure 1.1, foundations with horizontal spreading of the superstructure
load are considered as shallow foundations whereas foundations with vertical load distribution are
classified as deep foundations. Therefore, spread footings, combined footings and raft foundations,
where concentrated forces are distributed laterally, are considered as shallow foundations. Similarly,
piles are the most commonly used type of foundations where vertical distribution of the load takes
place.
Figure 1.1 - Classification of the foundation based on the load transfer mechanism.
Classification of piles
The British Standard Code of Practice for Foundations (BS 8004) places in three categories. These are
as follows.
Large displacement piles - comprise of solid-section piles or hollow-section piles with a closed end,
which are driven or jacked into the ground and thus displace the soil. All types of driven, and driven
and cast-in-place piles come into this category.
Small-displacement piles are also driven or jacked into the ground but have a relatively small cross-
sectional area. They include rolled steel H- or I-sections, and pipe or box sections driven with an open
end such that the soil enters the hollow section. Where these pile types plug with soil during driving
they become large displacement types.
F
F
(b)
Horizontal distribution
of the force
Vertical
distribution
of the force
(a)

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Replacement piles are formed by first removing the soil by boring using a wide range of drilling
techniques. Concrete may be placed into an unlined or lined hole, or the lining may be withdrawn as
the concrete is placed. Performed elements of timber, concrete, or steel may be placed in drilled holes.
Types of piles in each of these categories can listed as follows.
Large displacement piles (driven types)
1. Timber (round or square section, jointed or continuous).
2. Precast concrete (solid or tubular section in continuous or jointed units).
3. Prestressed concrete (solid or tubular section).
4. Steel tube (driven with closed end).
5. Steel box (driven with closed end).
6. Fluted and tapered steel tube.
7. Jacked-down steel tube with closed end.
8. Jacked-down solid concrete cylinder.
Large displacement piles (driven and cast-in-place types)
1. Steel tube driven and withdrawn after placing concrete.
2. Precast concrete shell filled with concrete.
3. Thin-walled steel shell driven by withdrawable mandrel and then filled with concrete.
Small-displacement piles
1. Precast concrete (tubular section driven with open end).
2. Prestressed concrete (tubular section driven with open end).
3. Steel H-section.
4. Steel tube section (driven with open end and soil removed as required).
5. Steel box section (driven with open end and soil removed as required).
Replacement piles
1. Concrete placed in hole drilled by rotary auger, baling, grabbing, airlift of reverse-circulation
methods (bored and cast-in-place).
2. Tubes placed in hole drilled as above and filled with concrete ass necessary,,
3. Precast concrete units placed in drilled hole.
4. Cement mortar or concrete injected into drilled~hole.:
5. Steel sections placed in drilled hole.
6. Steel tube drilled down.
Composite piles
Numerous types of piles of composite construction may be formed by combining units in each of
the above categories, or by adopting combinations of piles in more than one category. Thus
composite piles of a displacement type can be formed by jointing a timber section to a precast
concrete section, or a precast concrete pile can have an H-section jointed to its lower extremity.
Composite piles consisting of more than one type can be formed by driving a steel or precast
concrete unit at the base of a drilled hole, or by driving a tube and then drilling out the soil and
extending the drill hole to form a bored and cast-in-place pile.

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In Sri Lanka mostly cast in-situ concrete piles or driven piles are used for pile foundations.
Selection of pile type
Bored piles are constructed by drilling a hole into the ground and filling the hole with concrete
after inserting a reinforcement cage. In comparison, driven piles are constructed by driving a
preformed pile into the ground through application of hammer blows or vibration to the top of
the pile. General factors that should be considered in selecting the type of the pile between
driven piles and bored & cast in-situ piles are discussed in this section while detailed
construction procedures of the two piling methods are discussed in subsequent sections of this
chapter.
Bored piles can be very effectively used when a hard layer is present at shallow depths. The
structural loads can be easily transferred to the hard layer and generally the cast in-situ bored
piles are designed as end bearing piles. Other main advantage of bored piles is its ability to
penetrate minor obstructions, such as boulders, which cannot be penetrated using driven piles.
Such obstructions are commonly found in residual formations in the form of unewathered core
stones commonly referred to as boulders. A driven pile, terminated on a boulder, has a lower
carrying capacity due to the possibility of pile slipping along the face of the boulder and hence,
undergoing large displacement. Therefore, in such situations, use of driven piles is not advisable.
It is very common to find thick weathered or fractured rock above the solid bedrock. There are
certain places, where weathering profile is variable and the rock head is steeply sloping. Under
such situations, it is very important to socket the pile into the bedrock so that the full end bearing
could be mobilized without slipping of the pile toe. Thick weathered zones can be penetrated
using cast in-situ boring techniques and the structural loads can be transferred to the underlying
strong solid rock formations. Compared to driven piles, the ability of the bored piles to penetrate
fractured rock is a tremendous advantage in such situations.
The length of a bored pile can be adjusted easily and in a variable soil or bedrock profile, it is a
definite advantage. Diameter can also be varied and if large diameter piles are used, additional
cost associated with the construction of the pile cap, connecting a group of driven piles, can be
eliminated. The noise and ground vibrations associated with the other driven piling
methodologies are greatly reduced and installation process can be carried out even in a highly
built up area without environmental concerns. Moreover ground heaving associated with
installation of large volume displacement piles is not present with replacement type piles such as
bored and cast in-situ piles. The structural design of the pile should be carried out only
considering the working stresses and reinforcement required for driving and handling stresses
are not needed as in driven piles.
Drilling and concreting is carried out at a certain depth below the ground surface, and in most
cases under a drilling mud. The contamination of the concrete with the drilling slurry, formation
of voids within the pile, necking due to flowing of the sides into the unlined bore, collapsing of
the sides are some of the difficulties associated with bored and cast piles. Moreover, improper
cleaning of the pile bottom can cause considerable reduction in the end bearing capacity of bored
and cast in-situ piles. Formation of a thick hardened layer of bentonite along the sides of the
drilled hole is possible if the bentonite slurry is kept in the borehole for a long period of time.
During concreting the inability of the rising concrete surface to remove the hardened bentonite
may reduce the skin friction along the shaft of bored piles. Therefore, a proper quality
controlling program during the installation process is essential to minimize the defects in the
bored and cast in-situ piles.
As a preformed pile is used, certain concerns associated with bored piles such as necking,
improper cleaning of the borehole, defects due to mixing of concrete and drilling mud etc. are

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not present with driven piles. When the amount of soil replaced during the installation of the pile
is considered, driven piles fall into small or large volume displacement type. If a pile with a
large volume such as a precast concrete pile is used, the amount of soil displaced during
installation of the pile is large and it is considered as a large volume displacement pile. Whereas
if a pile with a relatively small volume such as a steel pile having a H or I section, is used as a
driven pile the amount of soil displaced during driving is small and such piles are classified as
small volume displacement piles. As a certain amount of soil is displaced during the installation
of a driven pile, the surrounding soil is compacted and thus the soil surrounding the pile is
improved giving rise to increase in the carrying capacity. In comparison, bored piles are
classified as replacement piles and the stress surrounding a borehole for a bored pile is relaxed
resulting reduction of the carrying capacity. Furthermore, the reduction of the skin frictional
capacity due to usage of drilling agents such as bentonite slurry is not present with driven piles.
The structural integrity and the capacity of driven piles are enhanced as the pile is formed under
controlled conditions above the ground surface. But the cast in-situ piles are formed below the
ground surface and in most situations under water making such high level of quality controlling
impossible. Another advantage of the driven piles is the ability to use the piling material
depending on the availability. As an example, when timber is available in abundant, timber piles
can be used for the foundation. Eventhough, noise and vibration generated during driving is very
critical in built up areas, such concerns are not significant in remote areas. Therefore, driven
piles can be used in such situations. If the noise is a concern special techniques such as silent
pile drivers can be used to reduce such environmentally unfriendly effects.

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Driven displacement piles
Timber piles
In many ways, timber is an ideal material for piling It has a high strength to weight ratio,
it is easy to handle., it is readily cut to length and trimmed after driving, and in favourable
conditions of exposure durable species have an almost indefinite, life. Timber piles used
in their most economical form consist of round untrimmed logs which are driven butt
uppermost. The practice of squaring the timber can be detrimental to its durability since it
removes the outer sapwood which is absorptive to creosote or some other liquid
preservative. The less absorptive heart-wood is thus exposed and instead of a pile being
encased by a thick layer of well-impregnated sapwood, there is only thin layer of treated
timber which can be penetrated by the hooks or slings used in handling the piles, or
stripped off by obstructions in the ground
Timber piles, when situated wholly below ground-water level, are resistant to fungal
decay and have an almost indefinite life. However, the portion above ground-water level
in a structure on land is liable to decay. Although creosote or other preservatives extend
the life of timber in damp or dry conditions they will not prolong its useful life
indefinitely. Therefore it is the usual practice to cut off timber piles just below the lowest
predicted ground-water level and to extend them above this level in concrete. If the
ground-water level is shallow the pile cap can be taken down below the water level.
Bark should be removed from round timbers where these arc to be treated with
preservative. If this is not done the bark reduces the depth of impregnation. Also the bark
should he removed from piles carrying uplift loads by skin friction in case it should
become detached from the trunk, thus causing the latter to slip. Bark need not be removed
from piles carrying compression loads or from fender piles of uncreosoted timber
(hardwoods are not treated because they will not absorb creosote or other liquid
preservatives).
The timber should be straight-grained and free from defects which could impair its
strength and durability. BS 8004 states that a deviation in straightness from the centre-
line of up to 25 mm on a 6 m chord is permitted for round logs but the centre-line of a
sawn timber pile must not deviate by more than 25 mm from a straight line throughout its
length. The Swedish Code SBS-S 23:6 (1968) permits a maximum deviation of 1% of
length between two arbitrarily selected measuring points which must be at least 3 m
apart.
The requirements of BS 8004 of the working stresses in timber piles merely state that
these should not exceed the green permissible stresses given in CP 112 for compression
parallel to the grain for the species and grade of timber being used. The Code suggests
that suitable material will be obtained from stress grades ss and better. Timber piles are
usually in a wet environment when the multiplying factors should be used to convert the
dry properties to the wet conditions. When circulating the working stress on a pile
allowance must he made for bending stresses due to eccentric and lateral loading and to
eccentricity caused by deviations in the straightness and inclination of a pile Allowance
must also be made for reduction in the cross-sectional area due to drilling or notching and
to the taper on a round log.

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This limitation is applied in order to avoid the risk of damage to a pile by driving it to
some arbitrary ‘set’ as required by a dynamic pile-driving formula and also to avoid a
high concentration of stress at the toe of a pile end bearing on a hard stratum. Damage to
pile during driving is most likely to occur at its head and toe.
The problems of splitting of the heads and unseen ‘brooming’ and splitting of the toes of
timber piles occur when it is necessary to penetrate layers of compact or cemented soils
to reach the desired founding level. This damage can also occur when attempts are made
to drive deeply into dense sands and gravels or into soils containing boulders, in order to
mobilize the required skin-frictional resistance for a given uplift or compressive load.
Judgment is required to assess the soil conditions at a site so as to decide whether or not it
is feasible to drive a timber pile to the depth required for a given load without damage, or
whether it is preferable to reduce the working load to a value which permits a shorter pile
to be used. As an alternative, jetting or pre-boring may be adopted to reduce the amount
of driving required. The temptation to continue hard driving in an attempts to achieve an
arbitrary set for compliance with some dynamic formula must be resisted. Cases have
occurred where the measured set achieved per blow has been due to the crushing and
brooming of the pile toe and not to the deeper penetration required to reach the bearing
stratum.
Damage to a pile can be minimized by reducing: as far as possible the number of hammer
blows necessary to achieve the desired penetration, and also by limiting the height of
drop of the hammer. This necessitates the use of a heavy hammer which should at least
be equal in weight to the weight of the pile for hard driving conditions, and to one-half of
the pile weight for easy driving. The German Code (DIN 183.04) limits the hammer drop
to 2.0 m normally and to 2.5 m exceptionally.
The lightness of timber pile can be an embarrassment when driving groups of piles
through soft clays or silts to a point bearing on rock. Frictional, resistance in the soft
materials can be very low for a few days after driving, and’ the effect of pore pressures
caused by driving adjacent piles in the group may cause the pile already driven to rise out
of the ground due to their own buoyancy relative to that of the soil. The only remedy is to
apply loads to the pile heads until all the piles in the area have been driven.
Heads of timber piles should be protected against splitting during driving by means of a
mild steel hoop slipped over the pile head or screwed to it. A squared pile toe can he
provided where piles are terminated in soft to moderately stiff clays. Where it is
necessary to drive them into dense or hard materials a cast steel point should be provided
(Figure 1.2). As an alternative to a hoop, a cast steel helmet can be fitted to the pile head
during driving. The helmet must be deeply recessed and tapered to permit it to fit well
down over the pile head, allowing space for the insertion of hardwood packing.

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Figure 1.2 - Typical shoes to be used with timber piles. (Technical specification EI
02G001, US Army Corp of Engineers)
Precast concrete piles
Precast concrete piles have their principal use in marine and river structures, i.e. in
situations where the use of driven-and-cast-in-situ piles is impracticable or uneconomical.
For land structures unjointed precast concrete piles are frequently more costly than
driven-and-cast-in-situ types for two main reasons.
1. Reinforcement must be provided in the precast concrete pile to withstand the
bending and tensile stresses which occur during handling and driving. Once
the pile is in the ground, and if mainly compressive loads are carried, the
majority of this steel is redundant.
2. The precast concrete pile is not readily cut down or extended to suit variations
in the level of the bearing stratum to which the piles are driven.
However, there are many situations for land structures where the precast concrete pile can
be more economical. Where large numbers of piles are to be installed on easy driving
conditions the savings in cost due to the rapidity of driving achieved may outweigh the
cost of the heavier reinforcing steel necessary. Reinforcement may be need in any case to
resist bending stresses due to lateral loads or tensile stresses from uplift loads. Where
high-capacity piles are to be driven to a hard stratum savings in the overall quantity of
concrete compared with cast-in-situ piles can be achieved since higher working stresses
can be used. Where piles are to be driven in sulphate-hearing ground or into aggressive
industrial waste materials, the provisions of sound high-qua1ity dense concrete is
ensured. The problem of varying the length of the pile can be overcome by adopting a
jointed type.
From the above remarks it can be seen that there is still quite a wide range of
employment for the precast concrete pile, particularly for projects where the costs of
establishing a precasting yard can be spread over a large number of piles. The piles can
be designed and manufactured in ordinary reinforced concrete, or in the form of pre-
tensioned or post-tensioned prestressed concrete members. The ordinary reinforced
concrete pile is likely to be preferred for a project requiring a fairly small number of

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piles, where the cost of establishing a production line for prestressing work on site is not
justifiable and where the site is too far from an established factory to allow the
economical transportation of prestressed units form the factory to the site. Precast
concrete piles in ordinary reinforced concrete are usually square or hexagonal and of
solid cross-section for units of short or moderate length, but for saving weight long piles
are usually manufactured with a hollow interior hexagonal, octagonal or circular sections.
The interiors of the piles can be filled with concrete after driving. This is necessary to
avoid bursting where piles are exposed to severe frost action. Alternatively drainage
holes can be provided to prevent water accumulating in the hollow interior. To avoid
excessive flexibility while handling and driving the usual maximum lengths of square
section piles and the range of working loads applicable to each size are shown in Table
1.1.
Where piles are designed to carry the applied loads mainly in end bearing, e.g., piles
driven through soft clays into medium-dense or dense sands, economies in concrete and
reductions in weight for handling can be achieved by providing the piles with an enlarge
toe.
Table 1.1 - Working loads and maximum lengths for ordinary precast concrete piles of
square section.
Pile size
(mm square)
Range of working Loads
(kN)
Maximum length
(m)
250 200 – 300 12
300 300 – 450 15
350 350 – 600 18
400 450 – 750 21
450 500 – 900 25
BS 8004 requires that piles should be designed to withstand the loads or stresses and to
meet other serviceability requirements during handling, pitching, driving and in service in
accordance with the current standard Code of Practice for the structural use of concrete.
If normal mixes are adopted a 40-grade concrete with a minimum 28-day cube strength of
40 N/mm2
is suitable for hard to very hard driving and for all marine construction. For
normal or easy driving, a 25-grade concrete is suitable. This concrete has a minimum 28-
day cube strength of 25 N/mm2
.
To comply with the requirements of BS 8110 precast piles of either ordinary or
prestressed concrete should have nominal cover to the reinforcement as follows.
Exposure conditions Normal cover for concrete grade of
25 30 40 50 and
over
Buried concrete and concrete continuously
under water
40 mm 30 mm 25 mm 20 mm
Alternative wetting and drying and freezing 50 mm 40 mm 30 mm 25 mm
Exposed to sea water and moorland water
with abrasion
__ __ 60 mm 50 mm

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Concrete cast in shell piles are constructed by driving a steel shell to a required depth by
using a mandrel and filling the shell with concrete after withdrawing the mandrel. Other
three types of concrete piles should be designed considering:
i. Bending stresses developed during handling;
ii. Dynamic stresses developed during driving; and
iii. Stresses due to working loads.
Longitudinal reinforcements are used to carry bending stresses developed during
handling of the precast concrete piles and lateral loads acting on the pile under working
condition. The bending moment diagram of single point handling and the corresponding
handling arrangement for minimum bending moment are given in Figure 1.3(b).
Similarly, the bending moment diagram and the arrangement for minimum bending
moment for double handling point mechanism is given in Figure 1.3(a). Steel stirrups are
used to carry driving stress acting on the pile. However, if the pile is subjected to static
vertical working loads, the reinforcement provided for handling and driving is mostly
redundant under working loads.
Figure 1.3 - Double and single lifting of precast piles: (a) Double lifting, bending moment
diagram and minimum bending moment; and (b) Single lifting, bending moment diagram
and the minimum bending moment.
Prestressed concrete piles have certain advantages over those of ordinary reinforced
concrete. Their principal advantage is in their higher strength to weight ratio, enabling
long slender units to be lifted and driven. However, slenderness is not always
advantageous since a large cross-sectional area may be needed to mobilize sufficient

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resistance in skin friction and end bearing. The second main advantage is the effect of the
prestressing in closing up cracks caused during handling and driving. This effect,
prestressed pile increased durability which is advantageous in marine structures and
corrosive soils.
The nominal mixes for precast reinforced concrete piles are related to the severity of
driving, and the working stresses appropriate to these mixes are shown in Table 2.6.
For economy in materials, prestressed concrete piles should be made with designed
concrete mixes with a minimum 28-day works cube strength of 40 N/mm2
. Metal shoes
are not required at the toes of precast concrete piles where they are driven though soft of
lose soils into dense sands and gravels or firm to stiff clays. A blunt pointed end appears
to be just as effective in achieving the desired penetration in these soils as a more sharply
pointed end and the blunt points is better for maintaining alignment during driving. A
cast-iron or cast-steel shoe fitted to a pointed toe may be used for penetrating rocks or for
splitting cemented soil layers.
During driving of the piles using an impact hammer, a compression stress wave travels
through the pile in the downward directions and reflected at the pile toe to travel upward
direction towards the pile top. If the end resistance at the pile toe is high (fixed end
condition) the reflected wave is compression and on the other hand, low resistance near
the pile toe results in tensile reflection at the pile toe. As a result, driving stresses are
maximum near the pile top and pile toe with reduced driving stresses in the middle
portion of the pile shaft. Therefore, more steel stirrups are provided near the pile top and
pile toe to take up the high driving stresses generated during driving. The requirements of
steel stirrups as specified in BS 8004 are given in Table 1.2 below.
Table 1.2 - The requirements of steel stirrups as specified in BS 8004 for driven precast
piles.
Volume of steel at head
and toe of pile
Volume of steel in
body of pile
Other requirements
0.6% gross volume over
distance of 3  pile width
from each end
0.2% of gross
volume spaced at
not more than ½ 
pile width
Lapping of shot bars with
main reinforcement to be
arranged to avoid sudden
discontinuity
Steel piles
Steel piles have the advantages of being robust, light to handle, capable of carrying high
compressive loads when driven on to a hard stratum, and capable of being driven hard to
a deep penetration to reach a bearing stratum or to develop a high skin-frictional
resistance, although their cost per metre run is high compared with precast concrete piles.
They can be designed as small displacement piles, which is advantageous in situations
where ground heave and lateral displacement must be avoided. They can be readily cut

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down and extended where the level of the bearing stratum varies; also the head of a pile
which buck1es during driving can be cut down an re-trimmed for further driving. They
have a good resilience and high resistance to buckling and bending forces. Types of steel
piles include plain tubes, box-sections, H-sections, and tapered and fluted tubes (Mono-
tubes). Hollow section-piles can be driven with open ends. If the base resistance must be
eliminated when driving hollow-section piles to a deep penetration, the soil within the
pile can be cleaned out by grabbing, by augers, by reverse water circulation drilling, or by
airlift. It is not always necessary to fill hollow-section piles with concrete. In normal
undisturbed soil conditions they should have an adequate resistance to corrosion during
the working life of a structure.
Where hollow-section piles are required to carry high compressive loads they may be
driven with a closed end to develop the necessary end-bearing resistance over the pile
base area. Where deep penetrations are required they may be driven with open ends and
with the interior of the pile closed by a stiffened steel plate bulkhead located at a
predetermined height above the toe. An aperture should be provided in the bulkhead for
the release of water, silt or soft clay trapped in the interior during driving. In some
circumstances the soil plug within the pile may itself develop the required base
resistance.
Concrete filling of light-gauge steel tubes is required after driving is completed because
the steel may be torn buckled or may suffer corrosion losses. Piles formed from thin steel
shells driven by means of an internal mandrel, which is withdrawn before filling the
shells with concrete.
The facility of extending steel piles for driving to depths greater than predicated from soil
investigation data has already been mentioned. The practice of welding-on additional
length of pile in the leaders of the piling frame is satisfactory for land structures where
the quality of welding may not be critical. A steel pile supported by the soil can continue
to carry high compressive loads even though the weld is partly fractured by driving
stresses. However, this practice is not desirable for marine structures where the weld
joining the extended pile may be above sea-bed level in a zone subjected to high lateral
forces and corrosive influences.
Bored and Cast In-situ concrete piles
Due to the presence of hard rock layers at relatively shallow depths, bored and cast in-situ
piles are very often used in Sri Lanka. Therefore, the construction procedure of bored and
cast in-situ piles are discussed here.
Replacement piles are installed by first removing the soil by a drilling process and then
construction the pile by placing concrete or some other structural element in the drilled
hole. As mentioned previously, bored piles are constructed by drilling a hole in the
ground and filling it with concrete with or without inserting a reinforcement cage. Since
the borehole in most cases is unlined, there is a possibility of flowing soft soils into the
borehole and forming a ‘necking’ in the pile shaft. Moreover, there could be collapsing of

124
the sides in cohesionless soils and fallen out debris may mix with fresh concrete resulting
in weak spots in the pile shaft. Furthermore, concreting is mostly done underwater
making it impossible to compact fresh concrete. Therefore, special construction
methodologies and precautions had to be followed to ensure a defect free sound bored
and cast in-situ pile. Compared with the construction of shallow foundations,
construction of deep foundation is a challenging task as the construction is carried out at
deeper levels without directly observing it. As a result, indirect quality control measures
should be adopted during the construction process.
In most sites, the ground water table is located at shallow depths and the top soil layers
contain cohesionless soils. Due to the loose soil conditions at the top levels of the ground,
the probability of collapsing of the ground is more. Therefore, it is very common to
install a casing of about 5 to 6m length at the top level of the borehole. If the ground
condition at the lower levels of the subsurface doesn’t contain very loose sandy material,
very often casing of the entire hole is not done and drilling is continued with filling the
hole with bentonite slurry. The cutting through the overburden is usually done by
auguring or chiseling and the cutting debris are removed from the hole using a bucket or
wash boring techniques. Drilling above the water table is usually done using an auger as
shown in Figure 1.4(a) and Figure 1.4(b) shows some auguring tools used for drilling.
(a)

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(b)
Figure 1.4 – Auguring above the water table: (a) Auguring during drilling; and (b) some
auguring tools used.
Figure 1.5 shows the installation of temporary casing during the drilling process for bored
and cast in-situ piles. After installation of the casing, the center of the casing is checked
as shown in Figure 1.6. During the drilling process, it is very common to use a drilling
fluid, such as bentonite slurry, to keep the sides of the borehole stable. The borehole is
filled with drilling fluid, when the borehole reaches the ground water table. For this
purpose, a bentonite reservoir is formed either surrounding the pile bore or away from the
pile bore location. Figure 1.7 shows a bentonite reservoir surrounding the pilebore and
Figure 1.8 shows a bentonite reservoir away from the pilebore location.
Figure 1.5 – Installation of a temporary casing.

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Figure 1.6 – Checking the location of the center of the pile.
Figure 1.7 – Bentonite reservoir surrounding the pile bore.

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Figure 1.8 – Bentonite reservoir away from the location of the pilebore.
Use of Bentonite as a drilling mud
Two types of natural bentonite exist: swelling bentonite which is also called sodium
bentonite and non-swelling bentonite or calcium bentonite. Sodium bentonite expands as
it can absorb several times its dry weight of water. It is mostly used in drilling mud in the
oil and gas well drilling industries as it exhibits low filter loss. However, non-swelling (or
low-swelling) bentonite has much higher filter or fluid loss than swelling sodium
bentonite and hence, it is not effective as a drilling fluid. As it is commonly accepted, the
drilling mud should perform or facilitate following tasks:
i. Remove cuttings produced by the bit at the bottom of the hole and carry them
to the surface;
ii. Lubricate and cool the drill bit during operation, as friction causes high
temperatures down-hole that can limit tool life and performance;
iii. Maintain hydrostatic equilibrium so that water from the surrounding soil do
not enter the borehole causing the wall to flow, kink and blow out. This is
achieved by adjusting the mud weight (density);
iv. Build a filter cake (or skin) on the wall of the drilled hole, preventing fluid
loss by mud invasion of penetrated formations; and
Pump used to circulate
bentonite

128
v. Support and prevent caving of the wall of the hole.
Typically if 3% or more of bentonite powder is dispersed in water, a viscous slurry is
formed which is thick when allowed to stand but becomes thin when agitated. This
phenomenon is referred to as thixotropy.
Bentonite slurry provides the stability to the borehole walls by two main actions: (i)
Formation of a filter skin termed “cake” at the interface of the slurry and the walls of the
excavated hole; and (ii) higher lateral pressure of the dense slurry pushing against the
filter skin and the walls of the excavated hole. The concreting of the hole should be done
in such a way to displace the slurry in the hole with the fresh concrete. However, if the
slurry full of hole is kept for a long period of time, a thicker and harder “cake” will be
formed on the internal walls of the borehole. If the soil surrounding the pile shaft is
permeable, the water in the bentonite slurry may seep into the surrounding area forming a
thicker filter cake. Some researchers have shown that it is possible to form a thin cake of
few millimeters even in clayey soils, which is quite impermeable. The formation of filter
cake in clayey soils to electrical forces or chemical reaction of bentonite suspension on
the wall of the borehole. It is argued that if the shear strength of the filter cake formed is
more than that of the fluid concrete, it cannot be scoured by the rising concrete surface
during concreting and may be left in place resulting in degradation of the development of
skin friction.
It is believed that the formation of the major portion of the filter cake, and hence the
reduction of the major portion of the skin friction capacity, takes place within first few
hours of the construction time and further increase in construction time have minor effect
on the reduction of the skin friction capacity. The formation of the filter cake, which
reduces the development of skin friction, takes place at a higher rate within first few
hours between the end of drilling and concreting. The delay time between the end of
drilling and concreting should be minimized to reduce the effects of the filter cake on the
development of skin friction in bored and cast in-situ piles. .
Formation of the filter cake takes place as the water in the bentonite slurry seeps to the
surrounding area in sandy soils or chemical action between bentonite slurry and the
surrounding clayey soils. Within the rock socketed length of the pile, seeping of the water
takes place through the cracks in the rock mass. If the cracks are open and filled with
high permeable debris, large quantity of water may seep into the surrounding area and
formation of the filter cake may be enhanced. However, if the rock mass surrounding the
socketed length is impervious, the filter cake formed may be limited to a very thin layer.
If the bottom of the borehole is cleaned immediately before concreting, there is a high
probability that the thin layer of the filter cake formed within the rock socket may be
scraped off.
Drilling below the water table
The drilling below the water table can be carried out using rotary drilling or percussion
drilling. If a rotary drilling method is used, a drilling bucket, as shown in Figure 1.9, is
used to remove material from the pile bore and it is emptied, as shown in Figure 1.10. On
the other hand, a chisel is lifted and dropped to loosen the material in the percussion
drilling and then wash boring is used to remove the debris from the pile bore. Figure 1.11

129
shows rigs used in the percussion drilling.
Figure 1.9 – Removing debris from the pile bore.
Figure 1.10 – emptying the drilling bucket.

130
Figure 1.11 – Percussion drilling rigs.
Figure 1.12 – Rotary drilling tool used to drill through rock.

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Termination of pile bore
During the drilling process for the piles, rotary drilling or percussion drilling techniques
are used to drill through the rock. Since the coring through the rock is very rarely done,
the residue coming out consists of very small rock particles, which hardly gives any
indication of the quality of the bedrock. Therefore, the quality of the bedrock should be
established by some other means.
It is not uncommon in Sri Lanka to find sites with fairly thick weathered rock layers
overlying the sound bedrock. In this type of sites, very often large variation in pile
lengths are reported within very short distances. Due to the steep dip angle of the bedrock
and the highly fractured nature of the bedrock, a pile termination criterion plays a special
significance in this type of sites. A detail site investigation including the investigation of
the bedrock is a must for these sites to design a suitable pile foundation and to plan the
construction phase of the foundation. Another weakness in the site investigation
procedure adopted in Sri Lanka is the lack of coordination between the site investigation
firm and the client and/or the consultant. If the site investigation firm informs the site
conditions, for example the variation of the bedrock profile and the quality of the bedrock
at the site, to the client and/or the consultant during the field investigation phase then, the
site investigation program can be modified to suit the site conditions.
If the establishment of bedrock was not properly done during the site investigation
process, identification of the bedrock in a variable bedrock profile is highly questionable.
In a site, where the bedrock elevation highly varies across the site, during the site
investigation stage rock drilling should be carried out at reasonable number of points
across the site to establish the bedrock level with a relatively high RQD. Such
investigation will not only give more data needed for the design of the pile foundation but
also will provide very important information needed for planning the construction
process as well.
In a typical site with varying bedrock profile, it is very difficult to identify the bedrock
and estimate the pile socketing length during drilling for the piles. Some piling
contractors use highly subjective criteria such as penetration rate of the drilling tool as a
guide to establish the bedrock level. However, socketing length and the termination
criterion of the piles based on the rate of penetration of drilling tools could be highly
misleading as the drilling through the bedrock could give high and low penetration rates
depending on the weathered nature of the fractures in the upper part of the bedrock and
the quality of the cutting tool.
The pile termination criterion, for a site with varying bedrock profile preferably should be
done after installation of a test pile near a location of a borehole used for field
investigation. The information obtained during drilling for the test pile and the load test
result obtained from the test pile should be used for determination of the carrying
capacity of piles in the site and in deciding the termination criterion to be used for
installation of the production piles. It should noted here that a test pile should be loaded
upto twice the working load as specified in section 6.2 of ICTAD/DEV/15, not upto 1.5
times the working load as testing of a working pile.
Once the termination criteria for the site are established, the drilling process should start
from one side of the site and proceed forward. Level of the bedrock should be marked on

132
a site map and preferably contours of the pile tip elevation should be plotted. The contour
map should be updated as the drilling progress and the pile tip elevation contours should
always be compared with the elevation of the bedrock established during the initial site
investigation process. Large variation of the pile tip elevation of a new pile from the
elevation shown by the contours should be carefully studied. For example, if the bedrock
is encountered at a higher elevation than the expected elevation of the bedrock from the
already established bedrock contours, drilling should be done to make sure that the pile
has not hit a core stone in the weathered rock layer.
Cleaning of the Borehole before Concreting
This is another very important aspect, which is not given due attention, during
construction of the piles. If the pile bottom is not properly cleaned before concreting is
done, there could be a layer of waste material present between the bottom of the pile and
the bedrock. As this material consists of unconsolidated loose debris, when the pile
bottom is loaded, it will undergo large settlement. The debris that is present may consist
of:
i. Granular material from the drilling operation through rock and soil, which is
in suspension with the drilling mud, may settle to the bottom of the borehole;
ii. Small block-like portions of soil and rock from the unlined wall of the
borehole may dislodge and fall down to the bottom of the borehole; and
iii. Ground water percolated through the pervious silty and sandy layers may
transport and deposit significant amount of sandy and silty material at the
bottom of the unlined borehole.
Even if the concreting procedure is methodical to give a defect free pile shaft, the
presence of the loose material below the pile bottom severely hamper the load carrying
capacity of the pile due to large settlement it undergoes.
Through the surveys carried out in Sri Lanka it is found that about 5% of the piles are not
according to the specifications and are categorized as ‘defective’. The analysis of the
load-settlement curves and the site conditions of the ‘defective’ piles indicated that the
piles have undergone large plunging type settlement under very small end bearing
resistance due to the presence of a relatively soft layer below the bottom of the pile.
Therefore, the most probable reason for presence of a soft layer beneath the pile toe is the
improper cleaning of the bottom of the borehole before concreting.
Concreting of Cast in-situ Bored Piles
The borehole may be dry, partially or completely filled with fluid before concreting.
Concreting under dry conditions should be done from dropping concrete from the ground
surface so that concrete ‘free falls’ onto the base of the borehole. A hoper or a guide
trunk should be used at the ground surface level to avoid contamination of the concrete
with soil near the ground surface level. The mix design of the concrete should be done to
produce a workable mix, which is self compacting without segregation. However,
segregation of fresh concrete may take place when the falling concrete hits the
reinforcement cage. To reduce the segregation of concrete this way, some contractors

133
sprinkle cement powder on to the reinforcement cage before concreting.
Concreting a borehole partially or completely filled with fluid is a difficult task and
requires careful planning and supervision to construct a defect free pile. Since it is not
possible to observe the actual concreting process taking place down the borehole, some
indirect quality controlling measures should be adopted to ensure defect free pile. The
contractor, in consultation with the consultant to the project, should device a suitable
quality control program prior to the beginning of the piling process. In devising such
quality control program, due consideration should be given to the subsurface conditions
of the site. It is very often observed that the piling contractors don’t change their
construction process to suite the subsurface condition.
The concreting of the pile under water should be carried out using a tremie pipe. The
tremie pipe should be watertight and the interior surface should be free from any
projections for unhindered passage of concrete through it. Typically 125mm to 200 mm
diameter tremie pipes are used to concrete bored piles in Sri Lanka. Usually larger
diameter pipes are used to concrete large diameter piles and/or concrete with large
aggregates. It should be reiterated here that before the commencement of concreting,
drilling mud at the bottom of the borehole should be checked for contamination.
The tremie pipe is assembled inside the borehole, which is full of bentonite slurry. The
funnel (or hoper) is attached at the top of the assembled tremie pipe, which is long
enough to reach the pile bottom as shown in Figure 1.13(a) A plug is placed at the bottom
of the hoper and a small volume of suitable buoyant material is placed between the
bottom of the fresh concrete in the funnel and slurry in the tremie pipe as shown in Figure
1.13(b). The purpose of the buoyant material is to keep the first batch of concrete mixing
with the slurry in the tremie pipe. Otherwise, during the falling of the first batch of
concrete through the tremie tube, washing of concrete and mixing it with bentonite slurry
may take place significantly weakening the concrete. The hopper is filled with concrete
with the removable plug placed at the bottom of the hopper. Then, the plug is jerked out
allowing fresh concrete to shoot down under its own weight to the bottom of the
borehole. Concrete rapidly moving down the tremie pipe may push the drilling mud in
the pipe through the bottom as shown in Figure 1.13(c). Thus the first charge of concrete
is placed and the bottom of the tremie is immersed in fresh concrete to create a sealed
environment inside the tremie from the drilling mud outside. There are two potential
problems associated with initial charging of tremie with concrete:
(i) Segregation of concrete during placement and;
(ii) Entrapment of air inside the tremie pipe.
To avoid these problems, the tremie should be filled slowly after placing the initial
charge. During the time period, from initial charging of the pile to end of concreting, the
bottom of the tremie pipe should be always kept below the top surface of the concrete
inside the borehole. The depth of embedment of the tremie pipe in the borehole should be
about 1.5m to 3.0m and higher depth of embedment should be used for concreting large
diameter piles.

134
Figure 1.13 - Concreting a borehole using tremie pipe: (a) Tremie is assembled in the
borehole; (b) A plug is placed at the bottom of the hopper and filled with concrete; (c)
Plug is removed and concrete moving through the tremie; and (d) Concreting continued
with bottom of tremie pipe immersed in fresh concrete.
The specific gravity of the drilling mud may go up with the degree of contamination of
the drilling mud with silt and other debris and the specific gravity of the drilling mud
should be less than 1.25 before the beginning of the concreting process. If the drilling
mud is contaminated with drilling debris and other substances, additional recycling or
substitution of the suspension is necessary so that the flow of fresh concrete can readily
replace the drilling mud at the bottom of the borehole. Concreting should be done in a
continuous operation without any interruptions. Therefore, the site engineer should make
necessary arrangements for continuous supply of concrete without delay. A contingency
plan should also in place to supply concrete if delay in the expected supply of concrete
happened due to some unforeseen reasons.
It is observed at most of the sites, that the tremie pipe is lifted up and lowered rapidly to
facilitate rapid flow of concrete. Since rapid lifting and lowering of the tremie causes the
mixing of drilling mud and the concrete within a certain zone surrounding the tremie,
such practice should be minimized or such movement should be limited to a small height.
Due to some reason if the tremie bottom is taken out of the fresh concrete, placement of
concrete should be stopped and the following procedure should be adopted in
recommencing the concreting process.

135
 The tremie should be gently lowered on to the surface of the
previously placed concrete with very little penetration. The tremie
should be filled with high slump concrete with higher cement content
and a new initial charging of the tremie should be done to displace the
laitance/scum at the top of the old concrete surface with fresh concrete.
 The tremie should be pushed further slowly making fresh concrete
sweep away laitance/scum in its way.
However, if there is any delay in recommencing the concreting of the borehole, the above
procedure may not be applied as replacement of laitance/scum of set or partially set
concrete cannot be effectively carried out. In such situations, a new pile fully or partially
replacing the problematic pile should be introduced.
Withdrawing the casing is another important process that has to be done during the
concreting process. The rate of withdrawing the casing is the governing factor. If casing
is withdrawn too fast, the
Minipiles and micropiles
Minipiles are defined in CIRIA report PGI(2.10)
as piles having a diameter of less than 300
mm, with working loads in the range of 50 to 500kN. The term “micro-pile is given to
those in the lower range of diameter. They can be installed by a variety of methods.
Some of these are:
i. Driving small-diameter steel tubes followed by injection of grout with or
without withdrawal of the tubes;
ii. Driving thin wall shells in steel or reinforced concrete which are Oiled with
concrete and left in place;
iii. Drilling holes by rotary auger, continuous flight auger, or percussion
equipment followed by placing a reinforcing cage and in-situ concrete in a
manner similar to conventional bore pile construction;
iv. Jacking-down steel tubes, steel box-sections. or precast concrete sections. The
sections may be jointed by sleeving or dowelling.
The principal use of minipiles is for installation in conditions of low headroom such as
underpinning work or for replacement of floors of buildings damaged by subsidence.
Factors governing choice of type of pile
The advantages and disadvantages of the various forms of pile described in 2.2 to 2.5
affect the choice of pile for any particular foundation project and these are summarized as
follows:
Driven displacement piles
Advantages
1. Material forming pile can be inspected for quality and soundness before driving.
2. Not liable to ‘squeezing’ or ‘necking’.
3. Construction operations not affected by ground water.

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4. Projection above ground level advantageous to marine structures.
5. Can be driven in very long lengths.
6. Can he designed to withstand high bending and tensile stresses.
Disadvantages
1. Unjointed types cannot readily be varied in length to suit varying level of bearing
stratum.
2. May break during driving, necessitating replacement piles.
3. May suffer unseen damage which reduces carrying capacity.
4. Uneconomical if cross-section is governed by stresses due to handling and driving
rather than by compressive, tensile, or bending stresses caused by working conditions.
5. Noise and vibration due to driving may be unacceptable.
6. Displacement of soil during driving may lift adjacent piles or damage adjacent
structures.
7. End enlargements, if provided, destroy or reduce skin friction over shaft length.
Driven-and-cast-in-place displacement piles
Advantages
1. Length can easily be adjusted to suit varying level of beating stratum
2. Driving tube driven with closed end to exclude ground water
3. Enlarged base possible
4. Formation of enlarged base does not destroy or reduce shaft skin friction
5. Material in pile not governed by handling or driving stresses
6. Noise and vibration can be reduced in some types by driving with internal drop-
hammer
Disadvantages
1. Concrete in shaft liable to be defective in soft squeezing soils or in conditions of
artesian water flow where withdrawable-tube types are used.
2. Concrete cannot be inspected after installation.
3. Length of some types limited by capacity of piling rig to pull out driving tube
4. Displacement may damage fresh concrete in adjacent piles or lift these piles or
damage adjacent structures. .
5. Noise and vibration due to driving may be unacceptable
6. Cannot be used in river or marine structures without special adaptation.
7. Cannot be driven with very large diameters.
8. End enlargements arc of limited size in dense or very stiff soils.
9. When light steel sleeves arc used in conjunction with withdrawable driving tube, skin
friction on shaft will be dest toyed or reduced.

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Bored-and-Cast-in-Place replacement piles
Advantages
1. Length can readily he varied to suit variation in level of bearing stratum.
2. Soil or rock removed during boring can be inspected comparison with site
investigation data
3. In-situ loading tests Can be made in large-diameter pile boreholes, or penetration tests
made in small boreholes.
4. Very large (up to 7.3m diameter) bases can be formed in favourable ground.
5. Drilling tools can break up boulders or other obstructions which cannot be penetrated
by any form of displacement pile.
6. Material forming pile is not governed by handling or driving stresses.
7. Can be installed in very long lengths.
8. Can be installed without appreciable noise or vibration.
9. No ground heave.
10. Can be installed in conditions of low headroom.
Disadvantages
1. Concrete in shaft liable to squeezing or necking in soft soils where conventional types
are used.
2. Special techniques needed for concreting in water-bearing soils.
3. Concrete cannot be inspected after installation.
4. Enlarged bases cannot be formed in cohesionless soils.
5. Cannot be extended above ground level without special adaptation.
6. Low end-bearing resistance in cohesionless soils due to loosening by conventional
drilling operations
7. Drilling a number of piles in group can cause loss of ground and settlement of
adjacent structures.

139
7.0 Design of Piles
Design criteria
Similar to a shallow foundation, failure of a structurally intact pile can be caused due to
two reasons: (i) shear failure of the soil surrounding the pile and; (ii) excessive settlement
of the foundation. Therefore, the task of the foundation designer is to find out an
economical pile to carry the working load with a low probability of shear failure while
keeping the resulting settlement to within allowable limits. In designing a single pile
against shear failure, it is customary to estimate the maximum load that can be applied to
a pile without causing shear failure, generally referred to as the ultimate carrying
capacity.
As in the case of shallow footings, two design approaches, (1) Allowable Stress Design
(ASD) method and (2) Load Resistance Factor Design (LRFD) method are available for
piles. The following sections will mostly elaborate the ASD method. The allowable stress
design (ASD) requires the following conditions:
Allowable loads
alln QFSP / (6.1 )
where
nP = ultimate resistance of pile
allQ = allowable design load
FS = factor of safety
The ultimate working load that can be applied to a given pile depends on the resistance
that the pile can produce in terms of side friction and point bearing (Figure 6.2). As the
pile is loaded at the pile top, the pile tends to move in the downward direction relative to
the surrounding soil. Therefore, the surrounding soil offers resistive force against that
relative movement. Hence the expression for the allowable load Pa on a pile would take
the following form:
FS
PP
Q
supu
all

 (6.3)
where
Ppu = ultimate point capacity
Psu = ultimate side friction
Determination of the ultimate carrying capacity of piles
There are mainly two different methods available to estimate the ultimate carrying
capacity of piles: i. static methods, and ii. dynamic methods.
Static methods can be further divided into following methods:
a. Using strength parameters of soil and/or rock;
b. Using empirical correlations and in-situ test results; and
c. Using static pile load test results.

140
Dynamic methods can also be sub-divided into following methods:
a. Using pile driving equations;
b. Using the wave equation method; and
c. Dynamic testing of the piles.
Figure 2.1 Load carrying mechanism of piles.
Static methods of estimation of the ultimate carrying capacity of piles
The ultimate carrying capacity of piles is the maximum load that can be applied on the
pile without causing shear failure of the surrounding soil both along the pile shaft and at
the pile bottom. As the skin friction may not be uniform along the pile shaft, the skin
friction is estimated by adding the skin friction along the pile shaft.
 siupu PPP ,
However, it is observed that the deformation required to develop the ultimate point
bearing capacity is much higher compared to the deformation required to develop the
ultimate skin frictional capacity. Therefore, some define the ultimate carrying capacity of
the pile as summation of the ultimate skin friction and the developed end bearing capacity
when the skin friction reaches the ultimate value.
 usipu PPP ,
The total pullout resistance of the pile may be estimated using the following Equation:
Friction along the pile
shaft (skin friction)
Resistance at the pile point

141
pusiu WPT   ,
uP = Ultimate (maximum) pile capacity in compression-usually defined as that load
producing a large penetration rate in a load test
Tu = Ultimate pullout capacity
Pp, u = ultimate pile tip capacity – seldom occurs simultaneously with ultimate skin
resistance capacity  usiP , : neglect for floating piles (which depends only on skin
resistance)
pP = tip capacity developed simultaneously with  usiP , : neglect for “floating piles”
 siP = skin resistance developed simultaneously with ultimate tip resistance Pp, u
: neglect for point bearing piles
 usiP , = ultimate skin resistance developing simultaneously with some tip
resistance Pp
Wp = weight of the pile being pulled
 = summation process over I soil layers making up the soil profile over length
of pile shaft embedment
The ultimate capacity of a pile can be generally written as:
bbusuu WPPP 
suP = ultimate shaft resistance
buP = ultimate base resistance
Wb = weight of the pile
Skin Friction
Development of skin friction in piles
A pile, which is in contact with the soil along its shaft, is loaded as shown in Figure
2.2(a). Due to the higher stiffness of the pile material relative to that of the soil, as the
load on the pile is applied, the pile tends to move in the downward direction relative to
the surrounding soil. This is similar to the situation, where two solid objects in contact
with each other, one of the objects tries to move relative to the other object. Naturally, a
resistive force is developed between the two objects to resist that attempted movement.
If two soil and pile elements in contact with each other are considered, the pile element
tends to move in the downward direction relative to the surrounding soil element as
shown in Figure 2.2(b). An imaginary space is created between the pile and the soil
elements, in reality they are in contact with each other. As the pile element tends to move

142
in the downward direction relative to the surrounding soil element, the soil element also
moves with it and the downward moving soil element applies an upward resistive force
(fs) on the pile element, as shown in Figure 2.2(b). The pile element apply an equal an
opposite downward force on the soil element.
As the downward displacement of the pile element increases, the resistive force
developed between the pile and the soil elements is increased as shown in Figure 2.2(c).
However, that resistive force cannot increase indefinitely. The resistive force developed
reaches a maximum, commonly referred to as the ultimate skin friction (fs,u). The
relationship between the skin frictional force and the downward deflection of the pile
element can be approximated as shown in Figure 2.2(c). The downward displacement
required to mobilize the ultimate skin friction resistance is relatively low and is in the
rage of 5 – 10mm. After the ultimate skin friction is mobilized, the pile and the soil
elements start to slip.
Figure 2.2 – Development of skin friction on pile.
Load Transfer Curves.
The axial force variation in the pile with the depth is referred to as the load transfer curve.
If the pile is not subjected to negative skin friction, the axial force in the pile is maximum
at the pile top and is equal to the applied force. Considering the static equilibrium of the
section of the pile upto a depth z, the following Equation could be written:

fs
fs
Pile element Soil element
Imaginary separation between
pile and soil elements
(a)
(b)
(c)
Relative displacement of
the pile element ()
Skin friction developed (fs)
fs,u
o

143
1sazat fPP 
From the above Equation it is clear that if the skin friction is acting in the upward
direction, the axial force decreases with the depth.
Considering the equilibrium of the small element of length dz shown in Figure, following
equilibrium equation could be written:
sdzzaaz dfPP   )(
Re-arranging the terms in the above Equation,
)( dzzaazs PPdf 
The skin friction at a given section is equal to the difference in the axial force at that
section.
Figure 2.2 - Axial force along the pile axis.
fs1
dfs
(a)
(b)
z
dz
Pa Pat
Paz
Pa(z+dz)

144
Figure 2.4 load transfer curves obtained by increasing the load acting on the top of a pile
in clayey soil.
Figure 2.5 load transfer curves obtained by increasing the load acting on the top of a pile
in sandy soil.

145
If the load transfer curve is vertical at a given section, it indicates that the skin friction in
that section is zero. Figures 2.4 and 2.5 are load transfer curves obtained by varying the
force acting at the top of piles and clay and sand respectively. Careful observation of
Figure 2.4 clearly indicates that when the force on the pile is increased from 300 kips, the
shape of the curves do not change significantly but the curves are shifted to the right (i.e.
the difference in the axial force between any two sections of the pile shaft does not
change significantly). This is due to reaching of the skin friction to the ultimate value.
Once the skin friction reaches the ultimate value along the entire pile shaft, the additional
load increased at the pile top directly increases the end bearing at the pile bottom. The
other point to note is the mobilization of the end bearing capacity. Initially only a very
small portion of the end bearing capacity is mobilized. But as the skin friction reaches the
ultimate capacity, the end bearing resistance increases significantly, finally reaching a
situation where load increment at the pile top is causing equal increment in the end
bearing resistance. From this it could be concluded that the initial load increments on
piles are taken up by mobilizing skin friction and very minimal end bearing is mobilized.
However, closer to failure load increments are entirely taken by increase in end bearing.
The load transfer curves shown in the Figure 2.5 also confirms the above facts and clearly
indicates that for HP 14 x 89 pile the shape of the load transfer curve do not change
significantly after 100 kips load at the pile top. This indicates that the skin friction has
reached the ultimate value by then. However, the end bearing capacity increases beyond
that. It was earlier mentioned that the skin friction reaches the ultimate capacity at low
deformation levels whereas the end bearing reaches the ultimate capacity at a very high
deformation levels.
Estimation of the ultimate skin frictional capacity of piles
Using soil strength parameters
As explained earlier, the development of the skin friction is due to the movement of the
pile shaft relative to the surrounding soil. At any level of the pile shaft a normal force (fn)
is applied on the pile shaft from the surrounding soil as shown in Figure 2.6.

146
Figure 2.6 – Development of frictional resistance.
Assuming that the soil element in contact with the pile is also solid, theory of friction
between two solid objects can be used to find out the frictional resistance developed.
From fundamental concepts in frictional resistance between two solid objects, the
frictional resistance (ffr) developed can be expressed as:
 tannnfr fff 
Where,
 - Coefficient of friction between the two objects; and
 - Friction angle between the two objects.
At a given location of the pile shaft, the normal force fn may be assumed to be a constant.
The friction angle, , is not a constant and it increases with the relative displacement
between the two objects. As the displacement of the pile element, as shown in Figure 2.2,
increases the frictional resistance, ffr, increases upto the ultimate value, fufr. The maximum
friction angle between the soil and the pile element is taken as a, which is generally a
function of the angle of internal friction of the soil. Therefore, the ultimate frictional
resistance (fufr) may be expressed as:
annufr fff  tan
Dividing both sides of the above Equation by the surface area of the element, As:
anufr  tan
Where
fs
fs
Pile element Soil element
Imaginary separation between
pile and soil elements
(a)
(b)
fn

147
ufr - Ultimate unit frictional resistance
n - Stress normal to the pile – soil interface
In addition to the frictional resistance developed between the pile and the soil, a unit
adhesive resistance, ca, may also be developed if the soil is cohesive. Therefore, the total
ultimate skin frictional resistance, us, may be expressed as:
aufrus c
By substitution,
anaus c  tan
It is generally observed that the:
'
vsn K  
Where Ks – coefficient of lateral earth pressure
v
/
- effective vertical stress at the level considered.
The lateral earth pressure coefficient is a function of the soil type, stress history and the
amount of disturbance caused to the surrounding by the pile installation process.
Thus,
asvaus Kc  tan
and
dzCP us
L
us 
0
 dzKcCP asva
L
us  tan
0
 
where
C = pile perimeter
L = length of pile shaft
Estimation of the skin friction in clayey soils
In clayey soils, the undrained condition is critical. Therefore, the ultimate skin frictional
resistance should be estimated using undrained strength parameters.
For clayey soils under undrained condition  = 0 and hence, a = 0. Therefore,
aus c
It was found that:
uus c 
Where
cu - Undrained cohesion of clay.
Different researches have suggested different relationships for  and one of the very
widely used simple relationship is given below:

148
Figure2.7 – Relationship between  and undrained strength (Su)
Example
Estimate the ultimate skin frictional resistance of the 400 mm x 400 mm, 9m long precast
concrete pile driven into clyey soil having undrained cohesion of 50 kPa.
Solution
From Figure 2.7,  = 0.9 (Using the relationship proposed by Bowles (1996))
509.0 xcuus   = 45 kPa
dzCP us
L
us 
0
cu = 50 kPa
400 mm x 400 mm driven
pile
9 m
Bowles (1996)

149
Since the us is constant with the depth, the total ultimate skin frictional capacity may be
estimated by multiplying us by the total surface area of the pile shaft.
Therefore,
  94.04.0245 xxxPus  = 648 kN.
Ultimate skin frictional capacity is 648 kN.
Exercise
Estimate the ultimate skin frictional capacity of the 9m long, 400 mm x 400 mm square
pile driven into a subsurface consisting of two clay layers having undrained cohesion of
25 kPa and 50 kPa as shown below. Estimate the ultimate skin frictional capacity of the
pile.
Estimation of the skin friction in clayey soils
The general skin friction capacity can be expressed by the Equation derived earlier:
asvaus Kc  tan
For sandy soils, ca = 0 and hence the above Equation is simplified to:
asvus K  tan
It is a very difficult task to accurately estimate the value of the lateral earth pressure
coefficient as it may depend on the type of soil, method of installation etc. As a result,
there are different methodologies proposed by various researches to estimate the
coefficient of lateral earth pressure closer to the pile. In this course, the -method
proposed by Burland (1972) is discussed.
-method
Burland (1972) made the following assumptions:
cu = 25 kPa
pile
9 m
cu = 50 kPa
4 m

150
1. Due to remolding adjacent to the pile, the effective stress cohesion intercept
reduced to zero;
2. The horizontal stress acting on the pile after dissipation of pore pressure is at least
equal to the horizontal stress prior to the installation of the pile (K0 condition); and
3. The major shear distortion during pile loading is confined to a relatively thin zone
around the pile shaft, and drainage of this thin zone either occurs rapidly during loading
or has already
occurred in the delay between driving and loading.
Considering the above assumptions, the skin friction can be expressed as:
aovus K  tan
Where
Ko - Coefficient of lateral earth pressure at rest and for normally consolidated
soils it is equal to (1 - sin).
Substituting Kotan a = , the ultimate skin frictional capacity is given by:
vus  
A particular attractive feature of this method is that if we use '
0 and1   SinK
then the range of  is from 0.27 to 0.30 in the pracital (range of 25°
to 45°
). This method
is more of then used with piles driven in cohesionless soil (when 0,0'
  ).
Variation of the effective overburden pressure (v) closer to the pile
In undisturbed ground, the effective overburden pressure increases with the depth, as the
weight of the soil above a certain level increases with the depth. However, this situation
changes closer to a pile when it is installed in the ground. Closer investigation of the skin
friction Equation given above reveals that at higher depths, if the effective overburden
pressure increases with the depth, the skin friction capacity should be very high. But
researches have found that the skin friction of piles do not increase without bounds with
the depth. The results of skin friction variation in sandy soils measured using model piles
are presented in Figure 2.8. It is clear from the variation of the skin friction with the
depth that the skin friction varies upto a certain depth but beyond that it remains constant.

151
Figure 2.8 – Variation of the skin friction with the depth (Vesic, 1967)
Based on the results of the research it is concluded that the effective vertical overburden
pressure closer to the pile is not similar to the vertical effective overburden pressure
under in-situ undisturbed conditions. The presence of the pile tends to change the stress
conditions closer to the pile. The pile provides some arching action and tends to reduce
the overburden pressure beyond a certain critical depth as shown in Figure 2.9.
Zc
W. T.
L
(a) (b)
wetdw+(Zc-dw)(sat- w)
wetdw+(L-Zc)(sat- w)

152
Figure 2.9 – (a) Vertical effective vertical stress distribution closer to the pile; and (b)
Vertical effective vertical stress distribution away from the pile.
It should be noted here that the skin friction developed on piles in sand varies with the
depth and the total resistance should be estimated considering such variations. The
relationship shown in Figure 2.10 can be used to determine the critical depth of a pile.
Figure 2.10 – Critical depth (zc) / pile diameter vs friction angle of the soil.
The angle of internal friction 1, prior to the installation of the pile, should be modified as
follows before using with Figure 2.10.
For driven piles;
10
4
3
1  o

For bored piles;
31  oo

Example I
A driven 400mm square, 9.0m long pile is installed in sandy soil layer having angle of
internal friction  = 32o
and cohesion c = 0. The water table is present at 1m below the
ground surface and the unit weight of soil above and below the water table are 16 kN/m3
and 17 kN/m3
respectively.
i. Determine the skin friction distribution along the pile; and
ii. Estimate the total ultimate skin frictional resistance on the pile.

153
Solution
i.
Equivalent diameter (d) of the pile can be determined by finding the diameter of the pile
having the same cross sectional area as the square pile.
mm
xB
d 451
42


Modification of the friction angle;
ooo
x 341032
4
3
10
4
3
1  
From Figure 2.10, Zc/d = 6, therefore, the critical depth Zc = 2.70m.
Effective overburden pressure at the water table level = 116x = 16 kPa
Effective overburden pressure at the critical depth level =   0.170.216  wsat 
= 28.2 kPa.
The angle adhesion, a = 0.75 = 24o
  21.0tansin1  a
Ultimate skin friction:
At the water table level = 1621.0 xv  = 3.36 kPa
At the critical depth level = 2.2821.0 xv  = 5.92 kPa.
Since the vertical effective overburden pressure closer to the pile remains constant
beyond the critical depth and  is also constant, the skin friction is constant below the
critical depth level.
wet= 16 kN/m3
Sat= 17 kN/m3
1 m
W. T
8 m
400 mm x 400 mm driven pile
 = 32o

154
The skin friction distribution can be graphically shown as below:
ii.
The total ultimate skin friction can be estimated by determining the area of the skin
friction variation with the depth.
Total skin friction upto the WT =  0.136.3
2
1
x = 1.68 kN
Total SF from 1.0 to 2.70m =   0.170.292.536.3
2
1
 = 7.79 kN
Total SF from 2.70 to 9.0m =  7.20.992.5 x = 37.3 kN
Total SF = 46.8 kN
Example II
A 800 mm diameter, 22 m long bored pile is installed through the subsurface shown in
the following Figure. The water table is present at 2 m below the ground surface and the
unit weight and the strength properties of the different layers are also given in the same
Figure.
i. Determine the skin friction distribution along the pile; and
ii. Estimate the total ultimate skin frictional resistance on the pile.
Zc=2.7m
16 kPa
28.2 kPa
28.2 kPa
3.36 kPa
5.92 kPa
5.92 kPa
Overburden pressure
variation closer to the pile
Ultimate skin friction
variation along the pile shaft

155
Solution
i.
The three layers present in the subsurface may be labeled from the top to bottom as L1,
L2 and L3 respectively.
Modification of the friction angle of the LI;
ooo
2933231  
From Figure 2.10, Zc/d = 5.5, therefore, the critical depth Zc = 4.4m. the critical depth is
within L1.
= 47.3 kPa.
Below the critical depth level, the effective overburden pressure closer to the pile is
constant with the depth.
Ultimate skin friction: For the bored and cast in-situ concrete piles, the angle of adhesion
a is assumed to be equal to the angle of friction of the soil.
The angle adhesion of L1, a =  = 32o
  29.0tansin1  a
  30.0tansin1  a
wet= 15 kN/m3
Sat= 17 kN/m3
2 m
W. T
7 m
800mm
diameter
bored pile
Medium dense
sand  = 32o
Dense sand  = 37o
and  = 18 kN/m3
Very dense weathered
rock layer  = 40o
and
 = 20 kN/m3
5 m
8 m

156
  30.0tansin1  a
Within L1:
At the water table level SF = 3029.0 xv  = 8.7 kPa
At the critical depth level SF = 3.4729.0 xv  = 13.7 kPa.
SF within L2 = 3.473.0 xv  = 14.2 kPa
SF within L3 = 3.473.0 xv  = 14.2 kPa
The skin friction distribution can be graphically shown as below:
ii.
The total ultimate skin friction can be estimated by determining the area of the skin
friction variation with the depth.
Total skin friction upto the WT =  0.27.8
2
1
x = 8.7 kN
Total SF from 2.0 to 4.4m =   0.24.47.137.8
2
1
 = 26.9 kN
Total SF from 4.4 to 9.0m =  4.40.97.13 x = 63.0 kN
Total SF from L2 =  0.82.14 x = 113.6 kN
Total SF from L3 =  0.52.14 x = 71.0 kN
Total SF = 283.2 kN
Zc=2.7m
30 kPa
47.3 kPa
47.3 kPa
8.7 kPa
13.7 kPa
Overburden pressure
variation closer to the pile
Ultimate skin friction
variation along the pile shaft
14.2 kPa
14.2 kPa
14.2 kPa

157
Determination of the end bearing capacity
The pile bottom is pressed against the soil beneath the pile toe and the bearing capacity
failure of the soil can occur. However, as the pile toe is at a greater depth below the
ground surface, the failure mode is normally the local shear failure and the failure pattern
is similar to the one shown in Figure 2.11.
Figure 2.11– Failure pattern below the pile toe.
As the failure pattern is different from the ones observed for shallow foundations, the
same bearing capacity equation, used for the estimation of the ultimate carrying capacity
of shallow foundations, may be used with modified bearing capacity factors.
BNNqcNq qcend
2
1


Where,
Nc, Nq, N = Bearing capacity factors
B = Width of the pile
q = Effective overburden pressure at the toe of the foundation
 = Unit weight of the material below the pile toe.
It is generally observed that the third term of the above bearing capacity Equation is
small compared to other two terms. Therefore, the third term of the above Equation is
neglected, if the width of the foundation is not large.
6 – 10 B
2 – 4 B

158
Skempton (1951) suggested the chart given in Figure 2.12 to obtain the bearing capacity
factor Nc.
Figure 2.12 – Bearing capacity factor Nc (Skempton, 1951)
It is evident from the chart given above that for a circular or square footing the maximum
value of the bearing capacity factor is 9.0 for L/Br ratio greater than about 4.0. The chart
given in Figure 2.13 is proposed by Berezantzev et al. (1961) for the estimation of the
bearing capacity factor N.
Figure 2.13 – Bearing capacity factor for N (Berezantzev et al., 1961)
The  value of the soil should be modified as below before using with the above chart.

159
For driven piles,
2
40'
1 



For bored piles, 3'
1  
Where '
1 = angle of internal friction prior to installation of pile
Example I
Estimate the ultimate end bearing capacity of the driven 400mm square pile driven 9m
into a clay layer having undrained cohesion 25 kPa.
For normally or slightly over consolidated soils, the undrained capacity is critical,
Therefore,  = 0 and Nq = 0.
cend cNq 
Since L/B > 4, Nc = 9.0
0.950xqend  = 450 kPa
Ultimate end bearing load, Pend = 450x0.4x0.4 = 72 kN.
Example II
Estimate the ultimate end bearing capacity of the 400 mm x 400mm driven pile shown in
the following diagram.
cu = 50 kPa
pile
9 m

160
Solution.
Equivalent diameter (d) of the pile can be determined by finding the diameter of the pile
having the same cross sectional area as the square pile.
mm
xB
d 451
42


Modification of the friction angle;
ooo
x 341032
4
3
10
4
3
1  
From Figure 2.10, Zc/d = 6, therefore, the critical depth Zc = 2.70m.
= 28.2 kPa.
The modified  angle to be used in Figure 2.13 =
2
40'
1 



= 36o
N = 90
902.28 xqNq qend  = 2538 kPa
Pend = 460 kN.
Exercise
Five boreholes are driven in a proposed building site to investigate the subsurface
condition for a 20-storey building. The subsurface at the site consists of loose silty sand,
stiff clay, completely weathered rock, and fractured rock. A typical subsurface condition
in a borehole and the estimated shear strength parameters are given in the following
wet= 16 kN/m3
Sat= 17 kN/m3
1 m
W. T
8 m
400 mm x 400 mm driven pile
 = 32o

161
Figure. As a trial design, 1000 mm diameter bored piles socketed 1m into the fractured
rock is considered. As a design engineer attached to the firm involved in the design,
Estimate the ultimate skin friction of a single pile upto the top surface of the weathered
rock layer.
Empirical correlations
There are large number of empirical correlations that can be used to estimate the skin
friction and end bearing of piles. However, these correlations should be used very
carefully as they are valid under the subsurface condition used to develop them.
Skin Friction
Correlations with the SPT blow counts:

162
Meyerhof (1956, 1976) proposed the following correlation for the estimation of skin
friction (fus):
55Nf mus  (kPa)
Where
m = 2.0 for piles with large volumes displacement
= 1.0 for small volume displacement piles
N55 = Statistical average of the blow count in the stratum.
Shioi and Fukui (1982) suggested the following empirical correlations for the estimation
of the skin frictional resistance.
For driven piles: 55,0.2 sus Nf  for sand; = 55,10 sN for clay (kPa)
For bored piles: 55,sus Nf  for sand; = 55,5 sN for clay (kPa)
Where
Ni,55 = Average blow count in the material indicated for the pile or pile segment length.
Correlations with the Cone Penetration Test (CPT)
Meyerhof (1956) and Thorburn and Mac Vicar (1971) suggested the following
relationship based on the CPT results:
cus qf 005.0 (kPa)
Where
qc = cone penetration resistance in kPa.
When the side friction (qs) of the cone is measured:
sus qf  (for small volume displacement piles) and;
sus qtof )0.25.1( (for large volume displacement piles)
End bearing
Correlations with the SPT blow counts:
Meyerhof (1956, 1976) proposed the following relationship for the estimation of the end
bearing capacity.
  N
B
L
Nq b
end 38040  (kPa)
Where
N = Statistical average of the SPT N55 numbers in a zone of about 8B above to

163
3B below the pile point.
B = Width or diameter of the pile.
Lb = Pile penetration depth into point-bearing stratum.
Correlations with the Cone Penetration Test (CPT)
Japanese use the following relationship to estimate the end bearing capacity:
cend qq  (in units of qc)
Where
qc = Statistical average of the SPT N55 numbers in a zone of about 8B above to
3B below the pile point.
Estimation of the ultimate carrying capacity from the pile driving formulae
This method is commonly used for the estimation of the ultimate carrying capacity of
driven piles. This method is based on two fundamental assumptions:
i. The pile is a rigid body with no elastic deformations; and
ii. The dynamic resistance of the pile is equal to the static resistance of the soil.
Most of the pile driving equations are based on the energy conservation during the
driving process and the equations of motion. Consider the hammer and the pile
immediately before the impact and after the impact shown in Figure 2.14.
Figure 2.14 – Hammer and pile velocities immediately before and after the impact.
u
up
W
Wp vp
v
Immediately before the
impact
Immediately after the
impact

164
The energy transfer from the hammer to the pile and the resulting deformation of the pile
can be diagrammatically shown as given in Figure 2.15.
Figure 2.15 – Energy transfer and the deformation of the pile during a single hammer
blow.
Considering the velocities of the pile and the hammer before and after the impact, and the
deformation of the pile during and after impact following relationships can be obtained.
g
Wv
WHeE f
2
2
1 
The efficiency of impact is
   
    1
2
22
22
22
22
E
E
vgwvgw
ugwugw
e
pp
pp
iv 



The law of impulse gives :
   pp
p
uv
g
W
uv
g
W

The coefficient of elastic restitution, n, is
p
p
vv
uu
n



Assuming 0pv , and eliminating u, up, and v,
p
p
iv
WW
WnW
e



2
The energy left after impact is











p
p
fivf
WW
WnW
WHeWHeeE
2
2
Various pile driving equations are developed by simplification of the above derived
Equation. Some of the commonly used pile driving Equations are given in the following
Table 2.1.

165
Table 2.1 Commonly used pile driving equations
Formula Equation for Ru Remarks
Senders
S
WH
Engineering
News CS
WH

C = 1.0 in. for drop hammer
0.1 in. for steam
hammer
0.1 WWp in. for
steam hammer on very
heavy piles
Eytelwein
(Dutch)
PWW
W
S
WH

.
Weisbach 2
2













l
SAE
L
WHAE
L
SAE PPP
Hiley
  P
Pf
WW
WnW
CCCS
WHe



2
321
.
2/1
See Tables 4.2, 4.3 and 4.4
for values of 321 ,,, CCCef ,
and n.
Janbu












S
WH
ku
1  dedu CCk  11
WWC Pd 15.075.0 
2
/ AESWHLe 
Danish
  2
1
/2 Pf
f
AEWHLeS
WHe

See Table 4.2 for ef values
Gates  SWHef 10log6.5 10
Units are inches and tons
(short)
 SWHef 25log0.4 10
Units are metric tons (1000
kg) and centimeters

166
Exercise I
Precast concrete piles with 350 mm x 350 mm cross sectional area and a length of 12m
are to be driven for the abutment of a bridge using a 2 ton hammer with a height of drop
of 1m. Estimate the termination set to be achieved if the working load on a pile is 400 kN
and a factor of safety 3 is required against ultimate failure, using Gates method.
Gates equation







S
WHeR fult
25
log0.4 10
Units are in metric tons (1000 kg) and centimeters. The symbols carry the usual
meanings.
Exercise II
The subsurface at a bridge site consists of a 2m thick recently placed fill followed by
5m thick normally consolidated clay layer, which is underlain by a thick hard
weathered rock layer. 12 m long precast concrete piles with 350 mm x 350 mm cross
sectional area are to be driven at this site for the bridge abutment using a 2 ton
hammer with a height of drop of 1m. Estimate the termination set to be achieved if
the working load on a pile is 400 kN and a factor of safety 3 of is required against
ultimate failure, using
a. Hiley method; and
b. Janbu’s method.
Hiley pile driving formula:


















p
pf
ult
WW
WnW
CCCS
WHe
P
2
321 2/)(
where
AE
LP
C ult
2
C1 = 3mm C3 = 2.5 mm
n = 0.4 ef = 0.75
W = Drop weight H = Drop height
L = Length of the pile A = Cross sectional area of the pile
Wp=Weight of the pile S = Set of the pile during driving
E = Young’s modulus of the pile material
Janbu’s formula:













S
WH
K
P
u
ult
1

167
Where









d
e
du
C
CK

1







W
W
C
p
d 15.075.0






 2
AES
WHL
e
(Assume unit weight of concrete and Young’s modulus of concrete as 24 kN/m3
and 27 x
106
kPa respectively)

168
3.0 Estimation of the settlement of a vertically loaded single pile
As any other foundation, the design of the pile foundations should be safe against
excessive settlements. Therefore, settlement of the pile should be estimated and checked
against the allowable settlement of the foundation. The settlement estimation methods
could be divided mainly into three types:
i. Methods involving empirical correlations;
ii. Semi – elastic approaches involving Load –transfer methods considering the
axial force at various points along the pile shaft;
iii. Methods based on theory of elasticity that involves the use of Midlin (1936)
equations for subsurface loading within the semi-infinite mass; and
iv. Use of the numerical methods such as Finite Element Method
Empirical correlations:
Meyerhof (1959) Method
Based on the field load test results on piles in sandy soils, Meyerhof suggested that the
settlement could be obtained from the Equation [3.1] if the applied load has a factor of
safety more than three against the applied load.
F
db
30
 [3.1]
Where
db - diameter of the pile base
F - Factor of safety on ultimate load (Must be > 3.0)
It should be noted here that there is no soil properties nor applied load is in the settlement
estimation equation and hence the validity of this method is highly questionable.
Focht (1976) method
Focht proposed an empirical equation to estimate the settlement of a pile using the
movement ratio, the ratio between the settlement of the pile () and settlement of the pile
acting as a column under the working load. Based on observation of piles in clayey soils
Focht suggested that the use of Equations [3.2] or [3.3] to estimate the settlement of a
single pile.
5.0
Col

if col > 8mm [3.2]
and
0.1
Col

if col < 8mm [3.3]

169
Methods based on theory of elasticity that involves the use of Mindlin (1936)
equations
Various researchers have used this approach to estimate the settlement of a single pile. In
most of these approaches, the pile is divided into a number of uniformly loaded elements,
and a solution is obtained by imposing compatibility between the displacements of the
pile and the adjacent soil for each element of the pile.
The displacement of the pile are obtained by considering the compressibility of the pile
under the axial loading. The soil displacements are obtained in most cases by using
Mindlin’s equations for the displacements within a soil mass caused by loading within
the mass. The difference between the various methods lies in the assumptions made
regarding the distribution of shear stress along the pile.
The method derived by Poulos and Davis (1968) is described below. The method
assumes a floating or frictional pile in a semi-infinite mass as shown in Figure 3.1.
Figure 3.1 – pile soil model used by Poulos and Davis (1968) for settlement estimation.
It is assumed in almost all the settlement analysis of piles that the pile and soil are stress-
free and that no residual stresses exists in the pile resulting from its installation. This
could be a false assumption for most of the practical situations. However, this error could
be somewhat minimized by selection of appropriate material properties.
If conditions at the pile-soil interface remain elastic and no slip occurs, the movement of
the pile and the soil should be equal. In the solution process only the vertical
displacement compatibility is considered and no lateral displacement of the pile is
considered.
P
D
L
h
Soil Young’s
modulus, Es,
and Poisson
ration, vs
Young’s
modulus of
the pile
material is Ep

170
The results of the analysis carried out by Poulos and Davis (1968) are presented in terms
of a parameter referred to as the relative stiffness (k) of pile. The relative stiffness factor
k is defined by Equation [3.4].
s
p
A
E
E
Rk  [3.4]
Where RA is the area ratio defined by Equation [3.5]
c
p
A
A
A
R  [3.5]
Where Ap – area of the pile cross section
Ac - Area bounded by the outer circumference of the pile.
Consider a pipe pile of outer diameter of Do and inner diameter of Di as shown in Figure
3.2. The relative are RA is given by Equation [3.6].
 
2
22
4
4
o
io
A
D
DD
R



 [3.6]
Figure 3.2 – Cross section of a pipe pile.
If the pile has a solid cross section without any cavities within it, the area ratio RA is
equal to unity.
Separation of the skin friction and end bearing capacities
The theory presented in Poulos and Davis (1996) can be used to determine the skin
friction distribution along the pile shaft and the hence to separate the skin friction and end
bearing.
A uniform floating pile in a semi-infinite elastic medium, the ratio between the skin
friction and the average skin friction for piles with K=5000 and K=50 are shown in
Figure 3.3. the variation shown in Figure 3 is obtained assuming no-slip condition
between the pile and the soil. It is clear from the Figure 3.3 that the stress distribution
becomes highly non-uniform, when the pile stiffness factor is smaller due larger
settlement of the pile near the top of the pile as a result of high compressibility of the
pile. However, as the pile stiffness becomes higher, the skin friction distribution becomes
more or less uniform. The Poisson ratio of the soil has a negligible effect on the skin
friction distribution.
Do
Di

171
Figure 3.3 – Stress distribution along the pile shaft of a floating pile.
If the elastic modulus of the bearing layer is Eb and the elastic modulus of the material
along the pile shaft is Es, the load transfer curves of end bearing piles, with different
Eb/Es, are shown in Figure 3.4.
Figure 3.4 – Variation of the axial force with the depth of the pile.
Based on the Mindlin (1936), Poulos and Davis (1996) suggested the following
methodology in estimation of the settlement of a single pile.

172
Settlement of a floating pile
According to Poulos and Davis (1996), the settlement of a single pile (ρ) may be
expressed as given in Equation [3.7].
DE
PI
s
 [3.7]
Where
P - Applied axial force
I - Settlement influence factor
Es - Elastic modulus of the surrounding material along the pile shaft
D - Diameter of the pile
Settlement influence factor (I)
Settlement influence factor I can be expressed as:
hvko RRRII  [3.8]
Where Io – Settlement influence factor for an incompressible pile (k=) in a semi infinite
elastic medium with a Poisson ratio =0.5. Io could be obtained from Figure 3.5.
Figure 3.5 – Settlement influence factor Io

173
Rk, Rv, and Rh are the modification factors, which could be obtained from Figures 3.6, 3.7
and 3.8 respectively.
Figure 3.6 – Modification factor Rk
Figure 3.7 – Modification factor Rv

174
Figure 3.8 – Modification factor Rh
Settlement of an end bearing pile
According to Poulos and Davis (1996), the settlement of a single pile (ρ) may be
expressed as given in Equation [3.9].
DE
PI
s
 [3.9]
Where
P - Applied axial force
I - Settlement influence factor
Es - Elastic modulus of the surrounding material along the pile shaft
D - Diameter of the pile
Settlement influence factor (I)
Settlement influence factor I can be expressed as:
bvko RRRII  [3.10]
Where Io – Settlement influence factor for an incompressible pile (k=) in a semi infinite
elastic medium with a Poisson ratio =0.5. Io could be obtained from Figure 3.5.
Rk, Rv, and Rb are the modification factors, which could be obtained from Figures 3.6, 3.7
and 3.9 respectively.

175
Figure 3.9 – Modification factor Rb
Estimation of the settlement of piles through layered medium
It is very rarely that the piles are installed through homogeneous medium. In reality, piles
are generally installed through layered soil mediums. Therefore, estimation of the piles
through layered medium should be performed. Figure 3.10 shows the settlement
influence factor (Io) estimated by various methods for a two layer medium with different
moduli ratio. The settlement influence factor (Io) estimated from more sophisticated
methods agree well with that estimated using the weighted average of the elastic moduli
of the two layers. Therefore, weighted average of the elastic moduli of the layered
medium is used in the estimation of the settlement of piles through layered medium as
shown in Figure 3.11 and Equation [3.11]. Similarly, the Poison ratio of the layered
medium is estimated using Equation [3.12].

176
Figure 3.10 – Settlement influence factor (Io) of a two layer medium.
Figure 3.11 – Layered medium
The elastic modulus to be used in the settlement estimation is given by Equation [3.11]
and that for the Poison ratio is given by the Equation [3.12]
hn
h3
h2
h1
Ei
E1 vi
E2 v2
E3 v3
Ei vi
En vn

177




 n
i
i
n
i
ii
av
h
hE
E
1
1
[3.11]




 n
i
i
n
i
ii
av
h
h
E
1
1

[3.12]
Example I
The thickness and elastic compressibility properties of the soil and rock layers near one
borehole are as follows:
Layer Thickness
(m)
Elastic modulus
(kPa)
Poisson ratio
Organic clay layer 7 2000 0.3
Medium dense sand
layer
8 15000 0.2
Completely
weathered rock
layer
13 50000 0.2
Highly fractured
rock
2 100000 0.2
Bedrock 1 200000 0.1
If a 800 mm diameter bored pile installed near this borehole is socketed 1m into the
bedrock layer, estimate the settlement of the pile under a working load of 2500 kN.
(Elastic modulus of concrete is 31.7 x 106
kPa)
Solution
31741
31
1000002135000081500072000



xxxx
Eavg
22.0
31
2.022.0132.0873.0



xxxx
Avg
1000
31741
107.31 6

x
E
E
k
Avg
p

178
hvKo RRRII 
Io = 0.05
Rk = 1.2
Rv = 0.89
Rb = 0.4
Settlement of a pile is
mm
x
xxxx
DE
PI
s
2
8.031741
)4.089.01.105.0(2500
1 
Exercise
If the drained compressibility parameters, given in Table, are assumed for the subsurface
layers and the bedrock shown in Figure, estimate the expected final settlement of a 600
mm bored pile installed upto the bedrock.
Table
Layer Drained Young’s
Modulus (kPa)
Poisson ratio (/) Thickness (m)
Fine sand
layer
10000 0.2 3
NC clay layer 15000 0.3 5
Weathered
rock layer
30000 0.2 6
Bed rock 150,000 0.1
Figure

180
7.0 Design of Pile Groups Subjected to Vertical Compressive Loads
Introduction
Depending on the carrying capacity of individual piles and the working load acting
through the structural elements, such as columns, there are situations that a single pile is
not capable of supporting the structural load. In such situations, it is customary to use a
group of piles to support such structural loads. Like any other type of foundations, the
pile group should also be designed considering:
i. Shear failure of the pile group – should have a reasonable factor of safety
against ultimate shear failure of the soil supporting the group; and
ii. The settlement of the group under the working loads – The settlement of the
pile group under the working load should be less than the allowable settlement
limit of the structure.
General configurations of pile groups are shown in Figure 4.1. When several piles are
clustered as shown in Figure 4.1, it is reasonable to expect that the soil pressures
produced from either side friction or point bearing will overlap as shown in Figure 4.2.
Figure 4.1 – Typical pile group patterns: (a) for isolated pile groups; and (b) for

181
foundation walls.
The resultant pressure intensity will depend on both the pile load and spacing. The
intensity of the stress due to overlapping, will obviously decrease with increased pile
spacing, s. However, it is not practical to have large spacing between piles as the cost of
the pile cap will be high with a larger pile cap. Therefore, center to center spacing (s)
between adjacent piles is a critical parameter in designing pile groups.
Figure 4.2 – Stressed zone surrounding an individual pile and the effect of the group
action.
Consider the stressed zones of an end bearing single pile and a pile group as shown in
Figure 4.3. The width and the depth of the stressed zone beneath a pile group is much
more than that of a single pile in the group. Due to the larger stressed zone beneath the
pile group, the group will undergo a higher settlement than the settlement of individual
piles under the same axial force. Therefore, settlement of a pile group may be critical
even though that of a single pile may be negligible.

182
Figure 4.3 – Extent of the compressed zones beneath: (a) single pile; and (b) Pile group.
Furthermore, due to the large stressed zone beneath the pile group, a weak layer present
may be incorporated in the settlement and ultimate carrying capacity estimation of a pile
group eventhough such layer will not be influencing the behavior of an individual pile as
shown in Figure 4.4.
Figure 4.4 – Effect of a weak soil layer beneath the pile tip: (a) Single pile; and (b) Pile
group.
Ultimate Carrying Capacity of a Pile Group
According to the load transfer mechanism, pile groups can be subdivided into two
categories as shown in Figure 4.5:
i. Free-standing pile group – the pile cap is not in contact with the soil; and
ii. Piled foundation – when the pile cap is in contact with the soil.
Relatively weak soil layer
Strong
soil
layer
Weak
soil
Layer
(a) (b)
Stressed zone

183
Figure 4.5 – Types of pile groups; (a) Piled Foundation; and (b) Free-standing pile group.
In estimation of the ultimate carrying capacity of pile groups, the main concern is the
capacity of the pile group in relation to the summation of the capacities of the individual
piles forming the group. Therefore, the group efficiency (Eg) is defined as given in
Equation [4.1].


 n
i
i
g
g
Q
Q
E
1
[4.1]
Where Qg - Ultimate capacity of the pile group
Qi - Ultimate capacity of the ith
pile in the group
n - Number of piles in the group.
Pile Groups in Clay
Free-standing pile group
There are mainly two methods of estimation of ultimate capacity of a free-standing pile
group in clay:
i. Using empirical correlations; and
ii. Method proposed by Terzaghi and Peck.
Following Example will be used to discuss the pile group in clayey soils.
Example 4.1
A pile group is formed by driving 9m long 400mmx 400mm precast concrete piles at
1125mm spacing as shown in Figure E4.1. As shown in Figure E4.1, the piles are driven
through a clay layer having cu = 25 kPa.
(a)
(b)

184
Figure E4.1.
It is needed to estimate the ultimate carrying capacity of a single pile in the pile group
given in Example 4.1.
Using the -method,
 = 0.95
Therefore, the ultimate skin frictional capacity (pult,skin) of a single pile is given by
Equation [E4.2].
360)944.0(250.1,  xxxxp skinult kPa [E4.2]
The end bearing capacity (pult,end) is given by Equation [E7.3].
36259)4.04.0(,  xxxp endult kPa [E4.3]
Therefore, the total ultimate capacity of a single pile is 396 kPa.
Equivalent diameter of a single pile is

44.04.0 xx
= 451 mm
Estimation of ultimate carrying capacity of a pile group in clay using empirical
correlations
Converse-Labarre formula
One of the empirical correlations very widely used earlier is the one referred to as the
Converse-Labarre formula. However, at present this method is not widely used.
According to the Converse-Labarre method, the efficiency of the pile group is given by
Equation [4.2].
9 m
Cu = 25 kPa
1125 mm
1125 mm
1125 mm 1125 mm
400 mm x 400
mm square
piles

185
   



 

mn
nmmn
Eg
11
90
1

[4.2]
Where
m - number of rows
n - Number of piles in a row
 - Tan-1
(d/s) in degrees
d - Pile diameter
s - center to center spacing between piles
Applying the Converse-Labarre formula for the pile group given in Example 4.1,
m = 3
n = 3
 = 





1125
4511
Tan = 21.9o
From the Converse-Labarre method,
Eg = 0.676
From equation [4.1],
9396676.0
1
xxQEQ
n
i
igg  
= 2409 kN
Feld’s rule
Calculated capacity of a pile is reduced by
16
1
for each adjacent pile, irrespective of the
pile spacing.
Applying the Feld’s rule for the pile group in Example 4.1:
Considering the plan view of the pile group, three types of piles are identifiable, as shown
in Figure E4.4, based on the number of adjacent piles to each type.
Figure E4.4 – Type of piles according to the number of adjacent piles.
Type A piles have 3 adjacent piles;
Type B piles have 5 adjacent piles; and
Type C pile has 8 adjacent piles.
A A
A A
B
B B
B
C

186
Therefore, the group capacity Qg can be estimated from Equation [E4.2].
386
16
1
814
16
1
514
16
1
31 xxxxxxQg 























 [E4.2]
25083869722.01  xxxnxQEQ gg kN.
Therefore, from the Feld’s rule, the efficiency of the pile group is 0.722.
The capacity of the pile group predicted by these empirical correlations vary widely and
there is no basis to select one method over the other method as there is no evidence to
support any method.
Terzaghi and Peck Method
This method is the most widely used method to estimate the carrying capacity of pile
groups. This method identifies two ways a pile group can fail: (i) failure of individual
piles in the group; and (ii) failure of the group as a one single block (or commonly
referred to as the block failure mode). Based on the above modes of failure, Terzaghi and
Peck defined the ultimate capacity as the lesser of : (a) The sum of capacities of the
individual piles in the group; or (ii) the bearing capacity of a block encompassing the pile
group as shown in Figure 4.6.
Figure 4.6 – Block failure of a pile group
L
Lg
Bg
Average Cohesion
along the pile
shaft, cavg
Average cohesion of the soil at the base is cb

187
Considering the pile group shown in Figure 4.6, the block failure capacity of the pile
group (Qblock) can be estimated from Equation [4.3].
  avgggcbggBlock LcLBNcLBQ  2 [4.3]
Group capacity given in Example E4.1 can be obtained using the Terzaghi and Peck
method.
Bg = 1125x2 + (400x2)/2 = 2650 mm
Lg = 2650 mm
L = 9 m
cb = cavg = 25 kPa
   25965.265.229.82565.265.2 xxxxQBlock 3947.5 kN.

9396
9
1
xQ
i
i 3564 kN
Therefore, the group capacity is 3564 kN and the failure mode is the individual pile
capacity mode.
Terzaghi and Peck method assumes that the failure takes place either due to individual
pile failure or due to block failure mode. Experimental results have shown the existence
of these two types of failure modes. For closer pile spacing, block failure becomes critical
whereas for wider spacing between piles in the group, individual pile failure becomes
critical. Terzaghi and Peck method assumes that there is a certain pile spacing below
which block failure takes place and for pile spacing higher than the critical spacing,
individual pile failure takes place. However, there is no experimental evidence, to support
such an abrupt transition from block failure to individual pile failure mode. In order to
obtain a more realistic estimate of the ultimate load capacity a group, the empirical
relationship given in Equation [4.4] could be used.
22
1
22
111
Blockg QQnQ
 [4.4]
Qg = Ultimate capacity of the group
Q1 = Ultimate capacity of a single pile
QBlock = Ultimate capacity of the block failure
n = Number of piles in the group
Pile Groups in Layered Medium
If the piles in the group are installed through layered soil medium, the portion of the piles
through the clay layers should be checked against block and individual failure. Individual
capacities of the piles should be considered for the portion of the piles through the sandy
soil layers. Consider the pile group through a layered soil medium as shown in Figure

188
4.7.
Figure 4.7 – Pile group through layered soil medium.
Bearing capacity factor for soil failure at the bottom of the block.
Pile Groups in Sandy Soils
There is ample evidence to believe that the ultimate carrying capacity of pile groups in
sandy soils is more than the summation of the ultimate capacities of individual piles in
the group. This is mainly due to the compaction of the sandy soils from the vibration
created by vibration generated due to pile driving. Figures 4.8 and 4.9 show the variation
of the group efficiency factor of pile groups in sand with the space-to-diameter ratio of
the pile group.
Sandy soil
(Skin friction of individual piles
through sand layer should be
considered)
Clayey soil
(Block failure mode should also be
considered through the clay layer)
Sandy soil
(Skin friction of individual piles
through sand layer should be
considered together with the
individual point bearing capacity)

189
Figure 4.8 – Variation of the group efficiency factor with the space-to-diameter ratio of
the piles in sandy soils.
.

190
Figure 4.9 – Variation of the group efficiency factor with the space-to-diameter ratio of
the piles in sandy soils (Vesic, 1969)
The broad conclusion drawn from the above data is that unless sand is very dense or the
piles are widely spaced, the overall efficiency is likely to be greater than 1. The
maximum efficiency is reached at a spacing of 2 to 3 diameters and generally ranges
between 1.3 and 2.

191
Settlement of Pile Groups
As shown in Figure 4.3, the influence zone below the pile bottom is very high compared
to that of a single pile. As a result the volume of soil compressed is more and hence, the
settlement of a pile group can be appreciable compared to a single pile even if the load
acing on a single pile in both cases is of the same order. Therefore, settlement analyses of
pile groups become important. Settlement estimation methods of pile groups can be
divided into following categories:
i. Empirical correlations;
ii. Equivalent raft method;
iii. Interaction between individual piles; and
iv. Finite Element Method.
Empirical Correlations
There are certain empirical correlations, developed by various researches based on their
experience, to estimate the settlement of pile group.
Skempton (1953) suggested the empirical relationship given in Equation [4.5] to estimate
the settlement of a pile group.
 
 2
2
1 12
94



B
BG


[4.5]
Where
G - Settlement of the pile group
1 - Settlement of a single pile under the average load
B - Width of the pile group in feet
For driven piles in sand, Meyerhof (1959) suggested the relationship given in Equation
[4.6] to estimate the settlement of pile groups.
2
1 1
1
3
5















r
s
s
G


[4.6]
However, there is very little experimental data to support these empirical methods and
these methods are not very often used at present.
Equivalent Raft Method
The principle behind this method is that the settlement of a pile group is equivalent to that
of a raft placed at a certain level in the subsurface. The level, at which the raft is placed,
is decided based on the subsurface condition and the load carrying mechanism of the
piles in the group. Some suggested levels for placing the raft are shown in Figure 4.10.

192
(a)
(b)
(c)
Figure 4.10 – Placement of an equivalent raft for settlement estimation: (a) friction piles
in soft soil; (b) piles through a soft layer end bearing on a hard granular material; and (c)
Pile group end bearing on bedrock.
Once the equivalent raft is placed and the load acting on the pile group is applied on the
raft, the problem becomes that of a shallow foundation. The settlement of the equivalent
raft may consist of immediate settlement and consolidation settlement components. If the
Equivalent raft
Soft clay
Bedrock
Equivalent raft
Distribution
of load at 1:4
L
2L/3
Soft clay
Dense granular soil
L
Equivalent raft
Distribution
of load at 1:4 2L/3
Soft clay

193
pile group is in clayey soils, consolidation settlement should be estimated in addtion to
the immediate settlement of the raft.
Estimation of the Immediate Settlement of the Group
Immediate settlement (i) at the center of the equivalent raft could be estimated from
Equation [4.9].
Fp
u
ni II
E
Bq 




 

2
1
2

 [4.7]
Where








 21
1
21
FFI p


[4.8]
qn - Net pressure from the equivalent raft
B - Width of the equivalent raft
Eu - Undrained Young’s modulus of the clayey soil
 - Undrained Poisson ratio of the clayey soil taken as 0.5
F1 & F2 - Factors from Figure 4.11
IF - Fox’s correction factor (Figure 4.12)
H - Thickness of the compressible material below the equivalent raft.
Figure 4.11 – Factors F1 and F2
H/B

194
Figure 4.12 – Fox’s depth correction factor IF.
It is extremely important to obtain a reliable value for the Young’s modulus of the soil
under the equivalent raft. Due to sample disturbance, unrealistically low Young’s
modulus values are obtained from stress-strain curves obtained from conventional
unconfined or triaxial compression tests in the laboratory. In-situ tests, such as plate load
tests done in boreholes or trial pits, may be used to obtain Young’s modulus, which
represent the field conditions much better. The stress-strain behavior of soil is highly
nonlinear. Therefore, the usual practice is to draw a secant AC to the stress-strain curve
corresponding to a compressive stress equal to the net foundation pressure at the base of
the equivalent raft, as shown in Figure 4.13.
Figure 4.13 – Estimation of the secant modulus of soil under the equivalent raft.
If the Young’s modulus of the soil under the equivalent raft is not constant with the
depth, Equation [4.9] cannot be used.

195
Development of negative skin friction
The load carrying capacity of piles come from skin friction and end bearing. The
development of the skin friction and the associated deformation of the soil and the pile
could be simply illustrated as shown in Figure 5.1 using a typical pile element with the
corresponding soil elements in contact with it. Figure 1(a) shows the equilibrium position
of the pile and the corresponding soil elements before loading the pile. Figure 5.1(b)
shows the location of the pile and soil elements after application of an axial force of P1.
(a) (b) (c)
Figure 5.1. Development of ultimate skin friction: (a) pile at rest condition with one
pile element in contact with the adjacent soil element; (ii) During loading
of the pile; and (c) slipping of the pile and soil elements after development
of the ultimate skin friction.
Figure 5.1(c) shows the location same elements after slipping between the pile and the
soil. When the slipping between the pile element and the soil element occurs, the
frictional resistance acting on the pile has reached the ultimate value (skin, ult). The shear
deformation required to mobilize the ultimate skin friction is considered to be relatively
small. Since the direction of the skin friction is opposite to the direction of the applied
axial load, the frictional resistance is referred as skin frictional resistance (or positive
friction). The skin friction is generated due to the relative deformation between the pile
and the soil elements.
If the soil surrounding the pile moves in the downwards direction relative to the pile, the
direction of the frictional force acting on the pile is in the downward direction. As a
result, the direction of the frictional force is same as that of the applied load. Therefore,
such frictional force acts as a load on the pile. This phenomenon is referred to as the
negative skin friction.
The surrounding soil can move in the downward direction as a result of the consolidation
settlement of the surrounding soil due to: (i) lowering of the ground water table; (ii)
surcharging of a clay layer as a result of placing a fill layer; and (iii) increased pore water
pressure in the vicinity of a driven pile. A portion of the pile must be fixed against the
vertical movement for the development of a significant negative skin friction on the pile.
If the pile moves with the consolidating soil, there is no negative skin friction developed.
In Sri Lanka, bored piles are end bearing and driven piles are driven to a strong bearing
Pile
element
skin
skin,ult
Pile
Soil elements
in contact
with the pile
element
P1 P2

196
layer. Therefore, significant negative skin friction can be developed. Most piling sites in
Sri Lanka with a soft surface soil layer are filled prior or after installation of the piles.
Such soil deposits, which are normally consolidated undergoes large amount of
consolidation settlement over a period of time after placing such a fill. The time taken for
the completion of the major portion of the settlement depends on the thickness and the
consolidation properties of the compressible soft soil layer. Another significant feature of
the ground condition in Sri Lanka is a presence of a relatively residual soil formation
generally consisting of thick weathered rock layer overlying the bedrock. Therefore, a
typical subsurface condition with a possibility of development of negative skin friction
consists of a fill layer underlain by a normally consolidated soft soil layer followed by a
residual weathered rock layer on the bedrock. A typical subsurface is shown in Figure
5.2.
Figure 5.2. Typical subsurface profile and development of negative and positive skin
friction on end bearing piles
The soil upto the bottom of the consolidating soft soil layer is moving in the downward
direction. However, the amount of downward movement is maximum at the ground
surface and zero at the bottom of the soft soil layer. If it is assumed that the pile is
restrained from moving at the bottom, the downward movement of the surrounding soil
generates a drag force at the top of the pile upto a certain depth where the relative
movement of the surrounding soil with respect to the pile is in the downward direction.
Beyond that depth, the movement of the surrounding soil is in the upward direction with
respect to the pile and, therefore, positive friction is developed on the pile. The plane
separating the positive and negative skin friction is referred to as the ‘neutral plane’.
However, if the pile is free to move at the bottom, the neutral point is also moving in the
downward direction and the neutral point is established at the location where the relative
downward movement of the surrounding soil with respect to the pile is zero. In such
situations, the drag force becomes a dragdown force and the pile settles under the
negative skin friction.
Design of piles against negative skin friction
It is very essential that the designer should investigate whether there is a possibility of
Residual soil
layer
Fill layer
Consolidating
soil l layer
Bedrock
Ground surface immediately
after placing the fill
Layer boundaries after
consolidation
Neutral plane

197
development of negative skin friction at the site concern. Presence of a soft layer or a
clay layer alone is not a reason for development of negative skin friction. Possibility of
placing a fill within the site area and lowering of the ground water table should be given
due consideration. If a cause for consolidation of the soft layer is not present or the soft
soil layer is sufficiently overconsolidated to prevent significant consolidation settlement
then the negative skin friction should not be considered in the design. It is evident from
the above discussion that an effective stress method such as -method should be used in
estimation of the negative skin friction acting on a pile. It was also discussed that the
negative skin friction acts only upto the neutral plane from the pile top and beyond that
positive friction is developed. However, it is very difficult to locate the neutral plane for a
multi-layer soil subsurface commonly encountered in Sri Lanka. Therefore, it is advisable
to estimate the negative skin friction upto the bottom of the consolidating soft layer using
an effective stress method such as -method, which gives the negative skin friction
(skin,neg):
/
, vnegskin  
Where /
v is the effective overburden pressure at the
Meyerhof (1976) proposed the skin friction factor () for driven piles given in Figure 5.3.
It should be noted here that the  value obtained from Figures 5.3 agrees well with the
experimental case studies present in this paper.
Figure 5.3 Negative skin friction factor  for piles driven into soft to firm clays,
Meyerhop (1976)
It should be noted here that considering the drag load as a load acting on the pile for
determination of the factor of safety could be conservative. Therefore, it is suggested that
the required ultimate carrying capacity of the pile be obtained by multiplying the working

198
load by a normal factor of safety. Once the carrying capacity of the pile is obtained,
neglecting the skin friction from the consolidating layer and other layers above it, it
should be divided by the working load plus the drag force to check whether a reasonable
factor of safety exists.

200
8.0 Pile Load Test
Testing of piles is done mainly due to the following reasons:
i. To evaluate the performance of pile at the preliminary or later stages in terms
of settlement and carrying capacity;
ii. To assess the structural integrity of the pile; and
iii. To obtain additional information required for the pile design such as: total
skin frictional capacity; distribution of the skin friction along the pile shaft
and mobilized end bearing.
Testing of the piles could be carried out at mainly two different stages:
i. Testing of test piles prior to the construction of working piles; and
ii. Testing of piles during construction stage of the working piles.
Piles tested during the construction stage could be further subdivided into (a) Piles for
preliminary testing and (b) Routine proof testing of piles.
The first type of testing is mainly conducted on test piles solely constructed to assess the
performance of the piles to get some inputs to the design and construction of the pile
foundation. The inputs that are generally obtained by testing test piles, includes design
strength parameters, compressibility of the pile and the suitability of a particular design
method. Since the construction and testing of the test pile(s) is very costly, the decision to
construct and test the test pile(s) should be done considering
 the importance and magnitude of the foundation,
 complexity of the subsurface condition at the site,
 possible influence of the installation process on the surrounding area, and
 the anticipated problems during the construction stage.
Number of test piles in a particular site may be decided mainly on the variability of the
subsurface condition within the site. The information obtained by testing the test pile(s)
may give the foundation designer the necessary confidence to go for an economical
design and construction methodology giving considerable savings to the client.
Moreover, the information gathered during the construction of test pile(s) could be used
very effectively to plan the construction process and the quality controlling program to be
adopted during the construction of the actual pile foundation. However, the results of the
test pile(s) should be considered together with the possible variability of the subsurface
and the bedrock condition within the site.
The second type of testing is mainly done during construction stage and they could be
further subdivided into
(a) Piles for preliminary testing and
(b) Routine proof testing of piles.
Preliminary test piles are constructed at the beginning of the construction stage to serve
as the test piles. The information intended to gather from a preliminary test pile are same
as that of a test pile. Since the preliminary test piles are constructed using the same
machinery and the methodology, they represent the working piles better than the test piles
constructed and tested prior to the construction of the working piles. However,
continuous construction of working piles after construction of the preliminary test piles

201
could be risky if the preliminary test piles prove that the design assumptions are not
correct. This happens as the cast in-situ bored piles cannot be tested until concrete gains a
reasonable strength, preferably until 28 days after casting, working piles constructed
during the time period between casting the test piles and testing them may not meet the
design requirements. Therefore, such piles should be supplemented with other piles or
replaced with new piles. The author has observed few such instances and found that the
cost of remedial work required is very high,
The impotence of following a well planned quality control program during construction
of bored piles was explained earlier. However, the quality of the constructed pile
foundation should be verified by subjecting selected piles to load tests. The number of
piles to be tested depends on many factors such as:
 importance and magnitude of the foundation,
 variability of the ground condition across the site,
 results of the quality control program, and
 factor of safety used in the deign.
Moreover, based on the other post construction testing such as, integrity testing of bored
piles, piles for load testing may be selected. Testing of bored and cast in-situ piles may be
broadly classified into: (i) load testing; and integrity testing.
Load Testing Piles
In load testing of piles, the pile is loaded upto a proof load or upto failure. Through load
testing it is aimed at establishing the load deflection behavior of the pile upto a proof load
or ultimate carrying capacity of the pile under the applied load. Depending on the type of
load applied on the pile, the load testing may be divided into following categories:
i. Compression load testing
ii. Tension load testing; and
iii. Lateral load testing.
Most of the bored piles constructed in Sri Lanka are subjected to compressive loads.
Therefore, most of the load testing carried out are compression load tests and only
compression load tests are considered in this lecture. However, there are situations, where
lateral load tests are carried out on bored piles to investigate the lateral load carrying
capacity, especially the piles, which are subjected to lateral earth pressure.
Compression load testing can be classified into three major types depending on the rate of
loading:
i. Static Load Testing;
ii. Dynamic load testing; and
iii. Statnamic load testing.
Static load testing can be further divided into two types: Conventional static load testing;
and static load testing using Osterburg cell.

202
Testing of bored piles is very often carried out to investigate the load-settlement behavior
of the tested pile upto a proof load. Normally in most cases, the pile is loaded upto about
1.5 times the working load to establish the load-settlement behavior. The termination of
the load testing at 1.5 times the working load is mainly due to the cost associated with
continuation of the loading beyond that level. If the pile doesn’t reach the ultimate failure
before the final test load, the ultimate carrying capacity of the pile cannot be obtained
directly from static load testing. Therefore, it is not possible to obtain the Factor of Safety
(FoS) of the pile against the shear failure of the soil surrounding the pile. Criteria for
satisfactory performance of the pile are specified based on the settlement of the pile at the
working load and the proof load. The specified settlement at the working load and the
proof load varies from testing standard used.
Advantages and Disadvantages of Different Methods of Load Testing
The information that can be gathered from load testing and the accuracy of the
information obtained vary from the type of the test used. More importantly, the cost of
testing depends very much on the method of the test adopted. Therefore, the selection of
the type of load test should be done based on the cost, information needed from testing
and the required accuracy of the information.
Conventional static load test produces the load – settlement curve as the final result of the
test. Normally the settlement at a given load and the likely plastic settlement
corresponding to an applied load could also be estimated, if unloading of the pile is also
done during testing. Separation of the total load into the skin friction and end bearing is
not directly available from a static load test result. Eventhough there are various methods
of separation of end bearing and skin friction such as method proposed by Van Weel
(1957) and Chin (1970), the accuracy of the predictions vary widely. If accurate skin
friction and end bearing information are needed, instrumentation of the pile with strain
gauges at different levels should be done. However, this is feasible only in the case of test
piles as the instrumentation of the pile should be done during installation. The load –
settlement curve obtained from static load test may not indicate the structural integrity of
the pile, if structural failure of the pile doesn’t take place at defective sections. Various
researchers have shown that the shape of the load-settlement curve can be interpreted to
get more information related to the tested pile. Thilakasiri (2007) attempted to describe
the reasons for commonly observed variations of the shape of the load-settlement curves
obtained by testing bored piles by using the Elastic Shortening Line (ESL). More detailed
discussion on the interpretation of the load – settlement curves obtained from static load
testing is presented in Chapter 6.
Static load test using Osterburg cell has the main advantage of obtaining the mobilized
skin friction and end bearing of the pile separately. Furthermore, load deflection curves
for the shaft and the end can be separately obtained. Since the Osterburg test doesn’t
require external reaction system, it can be performed under low headroom conditions.
However, the maximum load the pile can be tested is twice the minimum of ultimate skin
friction or the ultimate end bearing.
In contrast to a traditional load test, dynamic load test using the Pile Driving Analyzer

203
(PDA) yields direct information regarding the skin friction distribution along the pile
shaft, mobilized end bearing capacity, structural integrity of the pile etc. in addition to the
static load - settlement curve of the pile. Since the static behavior of the pile is estimated
from dynamic loading, the accuracy of the results of the dynamic load test depends on
number of factors such as the accuracy of the data gathered, experience of the person
analyzing the PDA data, data analyzing model used etc. Therefore, it is a common
practice to load test a certain number of piles using both static and dynamic methods and
calibrate the parameters used in interpretation of the PDA data using the load – settlement
curve obtained from the static load test.
Following facts should be considered in comparing the load – settlement curves obtained
from static and dynamic load tests.
i. Certain amount of creep settlement is included in the static load settlement
curve as the load is kept constant for a certain specified time period whereas
during dynamic load testing immediate response of the pile is obtained.
ii. Settlement measurement during static load testing is done using three to four
dial gauges and the average is taken as the settlement, which may be
erroneous if the difference between the dial gauge readings is high.
Various correlation studies were done throughout the world to investigate the accuracy of
the dynamic load tests. Most of the correlation studies were done on driven piles as the
PDA was originally developed to test driven piles. Most of the studies to compare static
and dynamic test results are done only considering the failure load predicted by the two
methods. Likins and Raushce (2002) showed that the Davisson failure criterion is the best
method to use in estimation of the failure load in the comparison studies.
Factors Affecting the Selection of the Load Testing Method for a given Site
Following factors should be considered in selecting the type of test to be used for load
testing of piles in a particular site.
i. The magnitude of the test load or the expected ultimate carrying capacity of
the pile – Depending on the magnitude of the test load the reaction load in the
static load test, input dynamic energy in the dynamic load test and the weight
of the cache system in the statnamic load test may vary. Economy of
providing the required loading method should be given due consideration in
selecting the type of the load test. Static load testing using the Osterburg Cell
test depends only on the relative magnitude of the ultimate skin friction or end
bearing, as the maximum load that can be applied on the pile during Osterburg
Cell test is twice the minimum of ultimate skin friction or ultimate end
bearing.
ii. The type of information that is expected to be gathered from load testing and
the level of accuracy of the required information – Certain test piles demands
more information from it than just the load – settlement curve. For example,
skin friction distribution along the pile shaft, mobilized end bearing, soil
stiffness of the surrounding soil, and integrity of the pile can only be obtained
through a dynamic load test. If more accurate such information is required,
use of static load test on an instrumented pile may be preferred.

204
iii. The environment within the site and surrounding it – Use of a static load test
to test piles in water, for example offshore piles, may be extremely difficult.
In such situations the use of the dynamic or statnamic load test may be
preferred. On the other hand if the space available is not sufficient to construct
a loading platform or reaction piles, performance of a static load test may not
be possible. On the other hand, if the ground vibration in a build up area is a
concern, dynamic load testing may not be possible.
iv. Relative cost involved with different types of load testing methods – There is a
significant difference between the costs required to conduct different type of
test. Moreover, preplanning is essential in conducting instrumented pile load
tests such as Osterburg cell test and static load test on instrumented piles.
Static Load Test
Static load testing is the conventional and most reliable method of load testing of piles. In
practice, load testing on piles are performed to either prove that the pile can safely bear a
design load or to establish a design load based on the ultimate pile bearing capacity
obtained from the test. In this test, load is applied at the top of the pile while monitoring
the settlement of the pile top. Based on the way the load is applied at the pile top, two
types of testing methods are followed, namely,
 Constant Rate of Penetration (CRP) test and
 Maintained Load (ML) test.
In the CRP test, compressive force on the pile is progressively increased to cause the pile
to penetrate the soil at a constant rate until failure occurs. According to BS 8004, the
penetration rate for friction piles in clay is 0.75mm/min whereas that for end bearing
piles in granular soil is 1.55mm/min. However, AS 2159 (1996) specifies penetration rate
for friction piles in clay as 0.5mm/min whereas that for piles in granular soil as 2.0
mm/min. AS 2159 states that the above rates of penetration could be halved or doubled
without significantly affecting the results of CRP tests. However, it should be emphasized
that the load test should be performed adhering to a certain testing specification and
commonly used testing specifications in Sri Lanka include BS 8004, ASTM D1143, and
AS 2159 etc. The testing specification to follow should be agreed upon when the
contractual agreement is reached.
The CRP method is essentially a test to determine the ultimate load on a pile and is
therefore, applied only to preliminary test piles or research type investigations. In the ML
test, load is increased in steps to some multiple, for example 1.5 times or twice the
working load, with the time-settlement curve recorded at each stage of loading and
unloading. At each loading step, the load is maintained constant until the rate of
settlement of the pile is smaller than a specific value, for example 0.25mm/hr. As it is
evident from the testing procedure, the CRP test could be finished within a short time
period than the ML test. Therefore, sometimes when the load-settlement behavior and the
ultimate carrying capacity are both needed, the pile could be tested using ML upto the
test load and thereafter continue loading to failure at a constant rate of penetration. A
study of the load settlement curves of both the ML and CRP methods show that the

205
ultimate load portion of each test agrees reasonably well. However, there are differences
prior to the ultimate region. This may be attributed to the ML test including greater soil
creep effect than the CRP test because of its longer duration.
As described previously, the test load applied on the pile should be decided based on the
factors such as possible negative friction and variation in the cutoff levels between the
test piles and working piles.
Conventional Static Load Testing of Bored Piles
Setting up of the Test
The static load is applied on the pile by jacking against a reaction system. The reaction
system is referred to as an arrangement of kentledge, piles, anchors or rafts that provides
the resistance against which load on the pile is applied. Kentledge and reaction piles are
the most widely used reaction systems. Pictures of two kentledges reaction systems are
shown in Figures 6.1(a) and 6.1(b) while that of a reaction pile system is shown in Figure
6.2. It may consist of precast concrete blocks, precast concrete piles, cast-iron blocks, or
any heavy dead weight. The kentledge is supported on a platform or grillage. The
platform or grillage is supported on the ground or on specially prepared foundations,
which are well clear of the test pile. Testing specifications put the limit on the minimum
distance between the test pile and the closest point of the foundation of the kentledge.
The foundation, supporting the kentledge, should be placed on stable ground and should
not give rise to any differential settlement. Figure 6.3(a) shows a situation where
kentledge had collapsed during a static load testing of a pile due to bearing failure of the
soil supporting the kentledge. The center of gravity of the kentledge is on the axis of the
test pile and the hydaulic jack should apply the load coaxial with the pile. Since the
potential energy stored in the kentledge under a large axial force is high, the kentledge
should be stable to avoid failure.
Figure 6.1 - Kentledge reaction systems used for static load testing of piles.

206
Figure 6.2 - Reaction system consisting of anchor piles and beam systems
(a) (b)
Figure 6.3 - Failure of Reaction systems (a) Failure of a kentledge; and (b) Failure of
tension bar system in reaction pile (Courtesy of Federation of Piling Specialists)
The total effective weight of the reaction system providing resistance to the hydraulic
jack should be at least 1.2 times the maximum test load intended to be applied on the pile.
If the weight of the reaction system is estimated from the volume and density of the
constituent materials, an adequate factor of safety against any variations should be
allowed. Failure of the reaction system could be disastrous and two occasions of failures
of kentledge reaction system and reaction pile systems are shown in figures 6.3(a) and
3.3(b) respectively.
The top of the pile should be cleaned to remove the weak contaminated concrete upto a
suitable depth below the cutoff level. If the elevation of the top level of the cleaned pile
head is low, it should be built up to the necessary height using a suitable grade of
concrete. Built up length of the pile should be of necessary strength to resist the axial
load at the pile top without undergoing excessive deformation. The built up section of the
pile should be coaxial with the original pile and the top surface of the pile should be
perpendicular to the axis of the pile. The arrangement shall be such that none of the test
load is carried by the ground, under the cap. The pile top should have enough area to
place the hydraulic jack and other deflection monitoring equipment. The top surface of
Anchor pile
Tested pile

207
the prepared pile head should be smooth and be free from any projections that might
cause stress concentration and crushing of concrete at the contact points.
The settlement monitoring devices such as mechanical dial gauges, and/or LVDT’s
should be supported on a reference beam, which is not affected by the deformation of the
soil during application of the load on the test pile. As the soil surrounding the pile shaft is
subjected to deformation during application of the axial force on the pile, the ground
supports of the reference beam must be placed away from the loaded pile, and the
foundations of the reaction system. The axis of the settlement monitoring devices should
be vertical to measure the vertical deflection of the pile and should possess an accuracy
specified by the testing specification. It is suggested that four dial gauges (or LVDTs)
must be placed on the pile head to monitor the settlement during load application and
placement of the dial gauges should be selected to catch differential settlement of the pile
head. The displacement measuring devise should have a sufficient travel to the maximum
expected differential settlement between the pile and the reference beam and should
possess a very low sensitivity against the possible environmental changes. The
deformations of the reference beam should be cross checked with respect to a nearby
fixed datum using suitably accurate survey equipment such as precise level. The entire
settlement monitoring system should not be subjected to any external deformations due to
nearby construction activities, vandalism or natural causes such as sunlight, wind etc.
In place of the reference beam and the displacement measuring devises, a wire suitably
tensioned between two stakes could be used to measure the deformation of the test pile.
The settlement is measured by the relative deflection between the wire and a graduated
scale fixed onto the pile.
The load is applied on the pile through a hydraulic jack placed between the pile head and
the reaction system. The jack used should have the required capacity to provide the test
load on the pile. The plunger of the jack should have enough run to undergo extension
equal to the downward deformation of the pile head and the upward deflection of the
reaction system. The total capacity of the jacks used to load the pile should exceed 20%
or more of the required maximum test load.
The load measuring devise may consist of a proving ring or load cell in addition to the
pressure gauge of the hydraulic jack. The load cell or the proving ring used to measure
the settlement of the pile head should have been calibrated recently and the certificate of
calibration performed by an approval agency must be produced. The readings of both the
load measuring device and the pressure gauge should be noted but only the reading of the
load measuring device should be used in interpretation of the measured load. The
pressure gauge readings are required for error checking of the measuring system. A
schematic diagram showing a typical pile load system is shown in Figure 6.4.

208
Figure 6.4 - Schematic diagram showing arrangement of the loading system and
measuring devises in a typical static load test
INTRPRETATION OF LOAD-SETTLEMENT BEHAVIOR OF PILES
The main final result obtained from the static load testing of piles is the load –settlement
behavior of the tested pile. The load – settlement curve thus obtained should be used to
estimate the carrying capacity of the pile. In estimation of the carrying capacity of a pile,
the possible failure of the pile due to serviceability failure of the pile, ultimate
geotechnical failure of the pile, or structural failure of the pile should be considered.
Serviceability failure is generally defined with respect to a certain threshold settlement
under loading and unloading of the pile. The threshold of the settlement specified by
varies depending on the standard considered and the type of testing method. AS 2159
(1995) specifies the acceptance criteria of the static compression load test as given in
Table 6.1
Table 6.1 Compression load test acceptance criteria (AS 2159 (1995))
Load Maximum deflection (mm)
Serviceability load 15*
and (after removing serviceability load ) 7*
1.5 x design action effect 50
And (after removing 1.5 x design action effects 30
*
Movement to include no more than 3mm creep over 5h (after load has been in place for
15 min.)

209
AS 2159 (1995) specifies the acceptance criteria given below when the pile is tested
using dynamic methods:
The pile head movement does not exceed:
(i) 12 mm at design serviceability load; and
(ii) 35 mm when the load is 1.5 x design action effect.
Design action effects mentioned in above may include other possible loading on the pile ,
such as: dragdown forces and additional force due to earthquake etc., during its working
life.
ICTAD/DEV/16 (1997) gives its performance specification for maintained pile load test
as given below:
For load cycle upto 1.0 x Working Load:
Maximum allowable gross settlement = 12 mm
Maximum allowable net settlement = 6 mm
For load cycle upto 1.5 x Working Load:
Maximum allowable gross settlement = 25 mm
Maximum allowable net settlement = 12 mm
The above serviceability criteria may vary depending on the type of structure supported
by the piles. For example, certain structures may tolerate much larger settlements than
what is specified in the above criteria. Even though such performance criteria of piles are
important for design purposes, they do not provide any indication of the capacity of the
pile.
The true ultimate failure of the pile is defined as the load corresponding to point in the
load – settlement curve, where settlement continues to increase without additional
increase in the load (point A in Figure 6.5). This condition may be very difficult to
achieve especially with end bearing bored piles socketed into the bedrock. There are
other definitions of ultimate capacity of piles such as:
1. The load beyond which there is an increase in gross settlement
disproportionate to the increase in load (point B in Figure 6.5).
2. The load indicated by the intersection of tangent lines drawn through the
initial, flatter portion of the gross settlement curve and the steeper portion of
the same curve (point C in Figure 6.5).

210
Figure 6.5 Typical load – settlement curve showing ultimate load based on some failure
criteria.
The location of points A and B depend on the personal judgment of the person
interpreting the load – settlement curve. The drawing of the initial tangent and the tangent
of the flatter portion of the curve also depends on the personal judgment. Moreover, the
scale of the graph might have a certain influence on the ultimate capacities determined by
these methods. A good method for estimation of the ultimate capacity should be
independent of the scale effects and the personal judgment of the interpreter.

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