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1991 William Barclay Parsons Fellowship
Parsons Brinckerhoff
Monograph 7
Seismic Design of Tunnels
A Simple State-of-the-Art Design Approach
Jaw-Nan (Joe) Wang, Ph.D., P.E.
Professional Associate
Parsons Brinckerhoff Quade & Douglas, Inc.
June 1993

First Printing 1993
Copyright © Jaw-Nan Wang and Parsons Brinckerhoff Inc.
All rights reserved. No part of this work covered by the copyright thereon may be
reproduced or used in any form or by any means — graphic, electronic, or mechanical,
including photocopying, recording, taping, or information storage or retrieval systems —
without permission of the publisher.
Published by
Parsons Brinckerhoff Inc.
One Penn Plaza
New York, New York

CONTENTS
Foreword ix
1.0 Introduction 1
1.1 Purpose 3
1.2 Scope of this Study 4
1.3 Background 4
Importance of Seismic Design 4
Seismic Design before the ‘90s 5
1.4 General Effects of Earthquakes 7
Ground Shaking 7
Ground Failure 8
1.5 Performance Record in Earthquakes 8
2.0 Seismic Design Philosophy for Tunnel Structures 13
2.1 Seismic Design vs. Conventional Design 15
2.2 Surface Structures vs. Underground Structures 15
Surface Structures 15
Underground Structures 16
Design and Analysis Approaches 16
2.3 Seismic Design Philosophies for Other Facilities 17
Bridges and Buildings 17
Nuclear Power Facilities 17
Port and Harbor Facilities 18
Oil and Gas Pipeline Systems 18
2.4 Proposed Seismic Design Philosophy for Tunnel Structures 19
Two-Level Design Criteria 19
Loading Criteria 20
3.0 Running Line Tunnel Design 25
3.1 Overview 27
3.2 Types of Deformations 27
Axial and Curvature Deformations 27
Ovaling or Racking Deformations 29
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3.3 Free-Field Axial and Curvature Deformations 31
Background 31
A Practical Approach to Describing Ground Behavior 31
Simplified Equations for Axial Strains and Curvature 33
3.4 Design Conforming to Free-Field Axial and Curvature Deformations 35
Background and Assumptions 35
Design Example 1: The Los Angeles Metro 35
Applicability of the Free-Field Deformation Approach 37
3.5 Tunnel-Ground Interaction 37
Simplified Interaction Equations 38
Design Example 2: A Linear Tunnel in Soft Ground 43
3.6 Special Considerations 48
Unstable Ground 48
Faulting 48
Abrupt Changes in Structural Stiffness or Ground Conditions 49
4.0 Ovaling Effect on Circular Tunnels 53
4.1 Ovaling Effect 55
4.2 Free-Field Shear Deformations 55
Simplified Equation for Shear Deformations 56
4.3 Lining Conforming to Free-Field Shear Deformations 58
4.4 Importance of Lining Stiffness 60
Compressibility and Flexibility Ratios 60
Example 1 61
Example 2 62
Summary and Conclusions 63
4.5 Lining-Ground Interaction 64
Closed Form Solutions 64
Numerical Analysis 76
Results and Recommendations 76
5.0 Racking Effect on Rectangular Tunnels 83
5.1 General 85
5.2 Racking Effect 86
5.3 Dynamic Earth Pressure Methods 87
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Mononobe-Okabe Method 87
Wood Method 87
Implications for Design 88
5.4 Free-Field Racking Deformation Method 88
San Francisco BART 90
Los Angeles Metro 90
Flexibility vs. Stiffness 90
Applicability of the Free-Field Racking Method 92
Examples 92
5.5 Tunnel-Ground Interaction Analysis 96
Factors Contributing to the Soil-Structure Interaction Effect 100
Method of Analysis 100
Flexibility Ratio for Rectangular Tunnels 102
Results of Analysis 112
5.6 Recommended Procedure: Simplified Frame Analysis Models 122
Step-by-Step Design Procedure 122
Verification of the Simplified Frame Models 128
5.7 Summary of Racking Design Approaches 133
6.0 Summary 135
Vulnerability of Tunnel Structures 137
Seismic Design Philosophy 137
Running Line Tunnel Design 138
Ovaling Effect on Circular Tunnels 139
Racking Effect on Rectangular Tunnels 139
References 141
iii

LIST OF FIGURES
Figure Title Page
1 Ground Response to Seismic Waves 6
2 Damage Statistics 11
3 Axial and Curvature Deformations 28
4 Ovaling and Racking Deformations 30
5 Geometry of a Sinusoidal Shear Wave Oblique to Axis of Tunnel 32
6 Sectional Forces Due to Curvature and Axial Deformations 39
7 Free-Field Shear Distortions of Ground Under Vertically Propagating
Shear Waves 57
8 Free-Field Shear Distortion of Ground (Non-Perforated Medium) 59
9 Shear Distortion of Perforated Ground (Cavity In-Place) 59
10 Lining Response Coefficient, K1 (Full-Slip Interface) 66
11 Lining Response Coefficient, K1 (Full-Slip Interface) 67
12 Lining Response (Thrust) Coefficient, K2 (No-Slip Interface) 69
15 Normalized Lining Deflection (Full-Slip Interface) 73
16 Normalized Lining Deflection (Full-Slip Interface) 74
17 Finite Difference Mesh (Pure Shear Condition) 75
18 Influence of Interface Condition on Bending Moment 78
19 Influence of Interface Condition on Lining Deflection 80
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20 Typical Free-Field Racking Deformation
Imposed on a Buried Rectangular Frame 89
21 Structure Stability for Buried Rectangular Frames 91
22 Soil-Structure System Analyzed in Example 93
23 Subsurface Shear Velocity Profiles 95
24 Free-Field Shear Deformations
(from Free-Field Site Response Analysis, SHAKE) 97
25 Structure Deformations vs. Free-Field Deformations, Case I
(from Soil/Structure Interaction Analysis, FLUSH) 98
26 Structure Deformations vs. Free-Field Deformations, Case ll
(from Soil/Structure Interaction Analysis, FLUSH) 99
27 Typical Finite Element Model
(for Structure Type 2) 103
28 Earthquake Accelerograms on Rock
West Coast 104
Northeast 105
29 Design Response Spectra on Rock
(West Coast Earthquake vs. Northeast Earthquake) 106
30 Types of Structure Geometry Used in the Study 107
31 Relative Stiffness Between Soil and a Rectangular Frame 108
32 Determination of Racking Stiffness 111
33 Normalized Racking Deflections
(for Cases 1 through 25) 115
34 Normalized Structure Deflections 116
v
Figure Title Page

36 Effect of Embedment Depth on Racking Response Coefficient, R 121
38 Simplified Frame Analysis Models 127
39 Moments at Roof-Wall Connections
Concentrated Force Model
40 Moments at Invert-Wall Connections
41 Moments at Roof-Wall Connections
Triangular Pressure Distribution Model
42 Moments at Invert-Wall Connections
vi
Figure Title Page

LIST OF TABLES
Table Title Page
1 Free-Field Ground Strains 34
2 Cases Analyzed by Finite Difference Modeling 77
3 Influence of Interface Conditions on Thrust 81
4 Cases Analyzed by Dynamic Finite Element Modeling 113
5 Cases Analyzed to Study the Effect of Burial Depth 120
6 Cases Analyzed to Study the Effect of Stiff Foundation 123
7 Seismic Racking Design Approaches 134
vii

ix
FOREWORD
For more than a century, Parsons Brinckerhoff (PB) has been instrumental in
advancing state-of-the-art design and construction of underground structures, and the
fields of seismic design and earthquake engineering are no exceptions. Almost three
decades ago PB’s engineers pioneered in these fields in the design and construction of
the San Francisco BART system, whose toughness during earthquakes, including the
recent Loma Prieta event, has been amply tested. Recently, PB developed state-of-the-art,
two-level seismic design philosophy in its ongoing Los Angeles Metro and Boston
Central Artery/Third Harbor Tunnel projects, taking into account both performance-level
and life-safety-level earthquakes.
This monograph represents PB’s continuous attempts in the seismic design and
construction of underground structures to:
• Improve our understanding of seismic response of underground structures
• Formulate a consistent and rational seismic design procedure
Chapter 1 gives general background information including a summary of earthquake
performance data for underground structures.
Chapter 2 presents the seismic design philosophy for tunnel structures and the
rationale behind this philosophy. Differences in seismic considerations between surface
structures and underground structures, and those between a seismic design and a static
design are also discussed.
Chapter 3 focuses on the seismic design considerations in the longitudinal direction of
the tunnels. Axial and curvature deformations are the main subjects. The free-field
deformation method and the methods accounting for tunnel-ground interaction effects are
reviewed for their applicability.
Chapter 4 takes a look at the ovaling effect on circular tunnel linings. Closed-form
solutions considering soil-lining interaction effects are formulated and presented in the
form of design charts to facilitate the design process.
Chapter 5 moves to the evaluation of racking effect on cut-and-cover rectangular
tunnels. This chapter starts with a review of various methods of analysis that are currently
in use, followed by a series of dynamic finite-element analyses to study the various factors
influencing the tunnel response. At the end, simplified frame analysis models are
proposed for this evaluation.
Chapter 6 ends this monograph with a general summary.

Acknowledgments
I wish to express my thanks to the Career Development Committee and Paul H.
Gilbert, the original initiator of the William Barclay Parsons Fellowship Program, for
selecting my proposal and providing continuous support and guidance throughout this
study. Thanks are also due to the Board of Directors of Parsons Brinckerhoff Inc. for
making the growth and flowering of an engineer’s idea possible.
The fruitful results of this exciting study would never have been possible without
technical guidance from three individuals — my fellowship mentors, Dr. George Munfakh
and Dr. Birger Schmidt, and the technical director of underground structures, Dr. James
E. Monsees. Their constant critiques and advice were sources of inspiration and
motivation.
Appreciation is due also to Tom Kuesel, who gave constructive technical comments
on the content of this study, and to Tim Smirnoff, who provided much of the tunnel
structural data of the LA Metro project. Ruchu Hsu and Rick Mayes deserve my thanks
for generously giving their time and comments on the draft of this monograph. Gratitude
is offered to many other individuals for numerous technical discussions on real world
seismic design issues for the ongoing Central Artery/Third Harbor Tunnel project and the
Portland Westside LRT project. They include: Louis Silano, Vince Tirolo, Anthony
Lancellotti, Dr. Sam Liao, Brian Brenner, Alexander Brudno, Mike Della Posta, Dr. Edward
Kavazanjian, Richard Wilson, and many others.
Very special thanks to Willa Garnick for her exquisite editing of the manuscript, and
to Randi Aronson who carefully proofread the final draft of the monograph. Their won-derful
work gave this fellowship study a beautiful finish. I also acknowledge the support
and contribution of personnel of the New York office Graphics Department, particularly
Pedro Silva who prepared the graphics and tables and laid out the text.
I simply could not put a period to this study without expressing thanks to my wife
Yvonne Yeh, my son Clinton and my daughter Jolene. Their sacrificing support of my
work through many late nights and weekends contributed the greatest part to this
monograph.
Jaw-Nan (Joe) Wang, Ph.D., P.E.
Professional Associate
Parsons Brinckerhoff Quade & Douglas, Inc.
June 1993
x

3
1.0 INTRODUCTION
1.1 Purpose
The purpose of this research study was to develop a rational and consistent seismic
design methodology for lined transportation tunnels that would also be applicable to other
underground lined structures with similar characteristics. The results presented in this
report provide data for simple and practical application of this methodology.
While the general public is often skeptical about the performance of underground
structures, tunnel designers know that underground structures are among the safest
shelters during earthquakes, based primarily on damage data reported in the past. Yet
one certainly would not want to run away from a well designed building into a buried tunnel
when seismic events occur if that tunnel had been built with no seismic considerations.
Most tunnel structures were designed and built, however, without regard to seismic
effects. In the past, seismic design of tunnel structures has received considerably less
attention than that of surface structures, perhaps because of the conception about the
safety of most underground structures cited above. In fact, a seismic design procedure
was incorporated into a tunnel project for the first time in the 1960s by PB engineers.
In recent years, however, the enhanced awareness of seismic hazards for
underground structures has prompted an increased understanding of factors influencing
the seismic behavior of underground structures. Despite this understanding, significant
disparity exists among engineers in design philosophy, loading criteria, and methods of
analysis.
Therefore, this study, geared to advance the state of the art in earthquake engineering
of transportation tunnels, has the following goals:
• To maintain a consistent seismic design philosophy and consistent design criteria
both for underground structures and other civil engineering facilities.
• To develop simple yet rational methods of analysis for evaluating earthquake effects
on underground structures. The methodology should be consistent for structures with
different section geometries.

1.2 Scope of this Study
The work performed to achieve these goals consisted of:
• A summary of observed earthquake effects on underground structures.
• A comparison of seismic design philosophies for underground structures and other
civil engineering facilities. Based on this comparison, seismic design criteria were
developed for underground tunnels.
• A quantitative description of ground behavior during traveling seismic waves. Various
modes of ground deformations and their engineering implications for tunnel design
are discussed.
• A review of current seismic design methodology for both circular mined tunnels and
cut-and-cover rectangular tunnels. Examples were used to study the applicability of
these conventionally used methods of analysis.
• The development of a refined (yet simple) method for evaluating the earthquake
ovaling effect on circular linings. This method considers the soil-structure interaction
effects and is built from a theory that is familiar to most mining/underground
engineers. To ease the design process, a series of design charts was developed,
and these theoretical results were further validated through a series of numerical
analyses.
• The development of a simplified frame analysis model for evaluating the earthquake
racking effect on cut-and-cover rectangular tunnels. During the process of this
development, an extensive study using dynamic finite-element, soil-structure
interaction analyses was conducted to cover a wide range of structural, geotechnical
and ground motion parameters. The purpose of these complex and time consuming
analyses was not to show the elegance of the mathematical computations. Rather,
these analyses were used to generate design data that could be readily incorporated
into the recommended simplified frame analysis model.
1.3 Background
Importance of Seismic Design
One of the significant aspects of the 1989 Loma Prieta earthquake in the San
Francisco area was its severe impact on the aboveground transportation system:
4

• The collapse of the I-880 viaduct claimed more than 40 lives.
• The direct damage costs to the transportation facilities alone totalled nearly $2 billion
(Werner and Taylor, 1990).
• The indirect losses were several times greater as a result of major disruptions of
transportation, particularly on the San Francisco-Oakland Bay Bridge and several
major segments of the Bay area highway system.
The San Francisco Bay Area Rapid Transit (BART) subway system was found to be
one of the safest places during the event, and it became the only direct public
transportation link between Oakland and San Francisco after the earthquake. Had BART
been damaged and rendered inoperative, the consequences and impact on the Bay area
would have been unthinkable.
The 60-mile BART system was unscathed by the earthquake because PB engineers
had the foresight 30 years ago to incorporate state-of-the-art seismic design criteria in their
plans for the subway tunnels (SFBARTD, 1960; Kuesel, 1969; and Douglas and Warshaw,
1971). The Loma Prieta earthquake proved the worth of their pioneering efforts.
Seismic Design Before the ‘90s
Based on the performance record, it is undoubtedly fair to say that underground
structures are less vulnerable to earthquakes than surface structures (Dowding and
Rozen, 1978; Rowe, 1992). Interestingly, some tunnels and shafts built without special
earthquake provisions have survived relatively strong earthquakes in the past — for
example, the Mexico City subway during the 1985 Mexico City earthquake. On the other
hand, some underground structures have been damaged severely in other events (see
Section 1.5).
Limited progress has been made in seismic design methodology for underground
tunnels since the work for BART, possibly because of favorable performance data, and
limited research work has been done toward a practical solution. The lack of a rational
methodology for engineers and the nonexistence of applicable codes has led to widely
varied measures taken by different engineers. For example:
• Some ignore seismic effects and fail to check the resistance of the structures to
earthquakes, even in highly seismic areas.
• Others conduct their seismic design for underground structures using the same
methodology developed for aboveground structures, without recognizing that
underground structures are constrained by the surrounding medium.
5

Figure 1.
Ground Response to Seismic Waves
(Source: Bolt, 1978)
6

7
Design based on such inappropriate measures may lead to the construction of unsafe
structures or structures that are too conservatively designed.
Although the progress of underground seismic design methodology is lagging, the
earthquake awareness in the country is not. Recent discoveries in seismology, geology
and geotechnical engineering have led to the belief that earthquake hazard is no longer
only a California problem. Many regions throughout the United States, Puerto Rico and the
Virgin Islands are now known to have the potential for tremors of similar or larger
magnitude than that of the Loma Prieta. This situation demands rethinking of the current
seismic design practice for our underground transportation systems.
1.4 General Effects of Earthquakes
In a broad sense, earthquake effects on underground tunnel structures can be
grouped into two categories – ground shaking and ground failure.
Ground Shaking
Ground shaking refers to the vibration of the ground produced by seismic waves
propagating through the earth’s crust. The area experiencing this shaking may cover
hundreds of square miles in the vicinity of the fault rupture. The intensity of the shaking
attenuates with distance from the fault rupture. Ground shaking motions are composed of
two different types of seismic waves, each with two subtypes. Figure 1 shows the ground
response due to the various types of seismic waves:
• Body waves travel within the earth’s material. They may be either longitudinal P waves
or transverse shear S waves and they can travel in any direction in the ground.
• Surface waves travel along the earth’s surface. They may be either Rayleigh waves or
Love waves.
As the ground is deformed by the traveling waves, any tunnel structure in the ground
will also be deformed. If the imposed deformation were the sole effect to be considered,
ductility and flexibility would probably be the only requirements for the design of tunnel
structures (from a structural standpoint). However, tunnel structures also must be
designed to carry other sustained loads and satisfy other functional requirements. A
proper and efficient tunnel structural design, therefore, must consider the structural
members’ capacity in terms of strength as well as ductility and flexibility of the overall
configuration.

8
Ground Failure
Ground failure broadly includes various types of ground instability such as faulting,
landslides, liquefaction, and tectonic uplift and subsidence. Each of these hazards may
be potentially catastrophic to tunnel structures, although the damages are usually
localized. Design of a tunnel structure against ground instability problems is often
possible, although the cost may be high. For example, it may be possible to remedy the
ground conditions against liquefaction and landslides with proper ground improvement
techniques and appropriate earth retaining measures.
It may not be economically or technically feasible, however, to build a tunnel to resist
potential faulting displacements. As suggested by Rowe (1992), the best solution to the
problem of putting a tunnel through an active fault is —- don’t. Avoidance of faults may
not always be possible, however, because a tunnel system may spread over a large area.
In highly seismic areas such as California, tunnels crossing faults may be inevitable in
some cases. The design approach to this situation is to accept the displacement, localize
the damage, and provide means to facilitate repairs (Kuesel, 1969).
1.5 Performance Record in Earthquakes
Information on the performance of underground openings during earthquakes is
relatively scarce, compared to information on the performance of surface structures, and
information on lined underground tunnels is even more scarce. Therefore, the summaries
of published data presented in this section may represent only a small fraction of the total
amount of data on underground structures. There may be many damage cases that went
unnoticed or unreported. However, there are undoubtedly even more unreported cases
where little or no damage occurred during earthquakes.
Dowding and Rozen (1978)
The authors reported 71 cases of tunnel response to earthquake motions. The main
characteristics of these case histories are as follows:
• These tunnels served as railway and water links with diameters ranging from 10 feet to
20 feet.
• Most of the tunnels were constructed in rock with variable rock mass quality.
• The construction methods and lining types of these tunnels varied widely. The
permanent ground supports ranged from no lining to timber, masonry brick, and
concrete linings.

Based on their study, Dowding and Rozen concluded, primarily for rock tunnels, that:
• Tunnels are much safer than aboveground structures for a given intensity of shaking.
• Tunnels deep in rock are safer than shallow tunnels.
• No damage was found in both lined and unlined tunnels at surface accelerations up to
0.19g.
• Minor damage consisting of cracking of brick or concrete or falling of loose stones
was observed in a few cases for surface accelerations above 0.25g and below 0.4g.
• No collapse was observed due to ground shaking effect alone up to a surface
acceleration of 0.5g.
• Severe but localized damage including total collapse may be expected when a tunnel
is subject to an abrupt displacement of an intersecting fault.
Owen and Scholl (1981)
These authors documented additional case histories to Dowding and Rozens’, for a
total of 127 case histories. These added case histories, in addition to rock tunnels,
included:
• Damage reports on cut-and-cover tunnels and culverts located in soil
• Data on underground mines, including shafts
The authors’ discussion of some of the damaged cut-and-cover structures is of
particular interest. These structures have the common features of shallow soil covers and
loose ground conditions:
• A cut-and-cover railroad tunnel with brick lining (two barrels, each approximately 20
feet wide) was destroyed by the 1906 San Francisco earthquakes. In this case, where
brick lining with no moment resistance was used, the tunnel structure collapsed.
• Five cases of cut-and-cover conduits and culverts with reinforced concrete linings
were damaged during the 1971 San Fernando earthquake. The damages
experienced by the linings included:
- The failure of longitudinal construction joints
- Development of longitudinal cracks and concrete spalling
9

10
- Formation of plastic hinges at the top and bottom of walls
The conclusions made by Owen and Scholl, based on their study, echoed the findings
by Dowding and Rozen discussed above. In addition, they suggested the following:
• Damage to cut-and-cover structures appeared to be caused mainly by the large
increase in the lateral forces from the surrounding soil backfill.
• Duration of strong seismic motion appeared to be an important factor contributing to
the severity of damage to underground structures. Damage initially inflicted by earth
movements, such as faulting and landslides, may be greatly increased by continued
reversal of stresses on already damaged sections.
Wang (1985)
In describing the performance of underground facilities during the magnitude 7.8
Tang-Shan earthquake of 1976, the author reported the following:
• An inclined tunnel passing through 13 feet of soil into limestone was found to have
cracks up to 2 cm wide on the side wall. The plain concrete floor heaved up 5 to 30 cm.
• Damage to underground facilities decreased exponentially with depth to 500 m.
Schmidt and Richardson (1989) attributed this phenomenon to two factors:
- The increasing competence of the soil/rock with depth
- The attenuation of ground shaking intensity with depth
Sharma and Judd (1991)
The authors extended Owen and Scholl’s work and collected qualitative data for 192
reported observations from 85 worldwide earthquake events. They correlated the
vulnerability of underground facilities with six factors: overburden cover, rock type
(including soil), peak ground acceleration, earthquake magnitude, epicentral distance,
and type of support. It must be pointed out that most of the data reported are for
earthquakes of magnitude equal to 7 or greater. Therefore, the damage percentage of the
reported data may appear to be astonishingly higher than one can normally conceive.
The results are summarized in the following paragraphs. Readers should be aware
that these statistical data are of a very qualitative nature. In many cases, the damage
statistics, when correlated with a certain parameter, may show a trend that violates an
engineer’s intuition. This may be attributable to the statistical dependency on other
parameters which may be more influential.

11
Figure 2.
Damage Statistics
(Source: Sharma and Judd, 1991)

• The effects of overburden depths on damage are shown in Figure 2A for 132 of the
192 cases. Apparently, the reported damage decreases with increasing overburden
depth.
• Figure 2B shows the damage distribution as a function of material type surrounding
the underground opening. In this figure, the data labeled “Rock (?)” were used for all
deep mines where details about the surrounding medium were not known. The data
indicate more damage for underground facilities constructed in soil than in competent
rock.
• The relationship between peak ground acceleration (PGA) and the number of
damaged cases are shown in Figure 2C.
- For PGA values less than 0.15g, only 20 out of 80 cases reported damage.
- For PGA values greater than 0.15g, there were 65 cases of reported damage out
of a total of 94 cases.
• Figure 2D summarizes the data for damage associated with earthquake magnitude.
The figure shows that more than half of the damage reports were for events that
exceeded magnitude M=7.
• The damage distribution according to the epicentral distance is presented in Figure
2E. As indicated, damage increases with decreasing epicentral distance, and tunnels
are most vulnerable when they are located within 25 to 50 km from the epicenter.
• Among the 192 cases, unlined openings account for 106 cases. Figure 2F shows the
statistical damage data for each type of support. There were only 33 cases of
concrete-lined openings including 24 openings lined with plain concrete and 9 cases
with reinforced concrete linings. Of the 33 cases, 7 were undamaged, 12 were
slightly damaged, 3 were moderately damaged, and 11 were heavily damaged.
It is interesting to note that, according to the statistical data shown in Figure 2F, the
proportion of damaged cases for the concrete and reinforced concrete lined tunnels
appears to be greater than that for the unlined cases. Sharma and Judd attributed
this phenomenon to the poor ground conditions that originally required the openings
to be lined. Richardson and Blejwas (1992) offered two other possible explanations:
-Damage in the form of cracking or spalling is easier to identify in lined openings
than in unlined cases.
-Lined openings are more likely to be classified as damaged because of their
high cost and importance.
12

2.0 SEISMIC DESIGN PHILOSOPHY
FOR TUNNEL STRUCTURES
13

2.0 SEISMIC DESIGN PHILOSOPHY
FOR TUNNEL STRUCTURES
2.1 Seismic Design vs. Conventional Design
The purpose of seismic design, like any civil engineering design, is to give the
structure the capacity to withstand the loads or displacements/deformations applied to it.
The philosophy employed in seismic design is different, however, from standard structural
engineering practice because:
• Seismic loads cannot be calculated accurately. Seismic loads are derived with a high
degree of uncertainty, unlike dead loads, live loads, or other effects such as
temperature changes. Any specified seismic effect has a risk (probability of
exceedance) associated with it.
• Seismic motions are transient and reversing (i.e., cyclic). The frequency or rate of
these cyclic actions is generally very high, ranging from less than one Hz to greater
than ten Hz.
• Seismic loads are superimposed on other permanent or frequently occurring loads.
Although seismic effects are transient and temporary, seismic design has to consider
the seismic effects given the presence of other sustained loads.
Conventional design procedure under permanent and frequently occurring loads calls
for the structure to remain undamaged (i.e., more or less within elastic range). Because of
the differences discussed above, however, proper seismic design criteria should consider
the nature and importance of the structure, cost implications, and risk assessment asso-ciated
with such factors as public safety, loss of function or service, and other indirect
losses (Nyman, et al, 1984).
2.2 Surface Structures vs. Underground Structures
For underground structures such as tunnels, the seismic design approach differs from
that of the surface structures (e.g., bridges and buildings).
Surface Structures
In the seismic design practice for bridges, the loads caused by an extreme event
(earthquake) in a seismically active region are often several times more severe than the
15

loads arising from other causes. To design a bridge to remain elastic and undamaged for
such infrequent loads is uneconomical and sometimes not possible (Buckle, et al, 1987).
Therefore, it is clearly not practical to use the same design approach to earthquakes as is
used for other types of loads. The seismic design philosophy developed for bridges
(AASHTO, 1991) is discussed briefly in Section 2.3.
16
Surface structures are not only directly subjected to the excitations of the ground, but
also experience amplification of the shaking motions depending on their own vibratory
characteristics. If the predominant vibratory frequency of the structures is similar to the
natural frequency of the ground motions, the structures are excited by resonant effects.
Underground Structures
In contrast, underground structures are constrained by the surrounding medium (soil
or rock). It is unlikely that they could move to any significant extent independently of the
medium or be subjected to vibration amplification. Compared to surface structures, which
are generally unsupported above their foundations, the underground structures can be
considered to display significantly greater degrees of redundancy thanks to the support
from the ground. These are the main factors contributing to the better earthquake
performance data for underground structures than their aboveground counterparts.
Design and Analysis Approaches
The different response characteristics of aboveground and underground structures
suggest different design and analysis approaches:
• Force Method for Surface Structures. For aboveground structures, the seismic loads
are largely expressed in terms of inertial forces. The traditional methods generally
involve the application of equivalent or pseudostatic forces in the analysis.
• Deformation Method for Underground Structures. The design and analysis for
underground structures should be based, however, on an approach that focuses on
the displacement/deformation aspects of the ground and the structures, because the
seismic response of underground structures is more sensitive to such earthquake
induced deformations.
The deformation method is the focus of this report.

17
2.3 Seismic Design Philosophies for Other Facilities
Bridges and Buildings
The design philosophy adopted in bridge and building codes (e.g., AASHTO and
UBC) is such that:
• For small to moderate earthquakes, structures are designed to remain elastic and
undamaged
• For more severe earthquakes, the intent is to avoid collapse but to accept that
structural damage will occur. This means that in a severe earthquake, the stresses
due to seismic loads will exceed the yield strength of some of the structural members
and inelastic deformations such as plastic hinges will develop (Buckle, et al, 1987).
Using this design philosophy for a severe earthquake, the structural members are
designed for seismic forces that are lower than those anticipated if the structures were to
remain elastic. This reduction in seismic forces is expressed by the response modification
factor in the codes. At the same time, these codes also require that catastrophic failures be
prevented by using good detailing practice to give the structures sufficient ductility.
Normally, the larger a response modification factor used in the design of a member, the
greater the ductility that should be incorporated in the design of this member. With this
ductility the structures are able to hang together, even when some of the members are
strained beyond their yield point.
Although the two-level design concept (small versus severe earthquake) is adopted in
the bridge and building codes, the explicit seismic design criteria specified in these codes
are based only on a single level of design earthquake — the severe earthquake. Typical
design shaking intensity specified in these codes (ATC, 1978; UBC, 1992; AASHTO, 1983
and 1991) is for an earthquake of about a 500-year return period, which can be translated
into an event with a probability of exceedance of about 10 percent during the next 50 years.
Nuclear Power Facilities
Two-level earthquake design philosophy is adopted for nuclear power facilities:
• For the Operating Basis Earthquake (OBE), the lower-level event, the allowable
stresses in all structural members and equipment should be within two-thirds of the
ultimate design values.
• For the Safe Shutdown Earthquake (SSE), the higher-level event, stresses caused by
seismic loads should not exceed the ultimate strength of the structures and
equipment.

Port and Harbor Facilities
Neither standard seismic codes nor universally accepted seismic design criteria exist
for waterfront facilities such as berthing (wharf) structures, retaining structures, and dikes.
Recent advances in seismic design practice for other facilities, however, have prompted
the development of several project specific seismic design criteria for waterfront facilities
in high seismic areas (POLA, 1991; Wittkop, 1991; Torseth, 1984).
The philosophy employed in the design, again, is based on two-level criteria:
• Under an Operating Level Earthquake (OLE), a smaller earthquake, the structures
should experience little to no damage and the deformations of wharf structures should
remain within the elastic range. Generally, the OLE is defined to have a probability of
exceedance of 50 percent in 50 years.
• Under a Contingency Level Earthquake (CLE), a larger earthquake, the structures
should respond in a manner that prevents collapse and major structural damage,
albeit allowing some structural and nonstructural damage. Damage that does occur
should be readily detectable and accessible for inspection and repair. Damage to
foundation elements below ground level should be prevented (POLA, 1991).
Generally, the CLE is to have a probability of exceedance of 10 percent in 50 years.
The risk level defined for the CLE is similar to that of the design earthquake adopted in
bridge and building design practice.
Oil and Gas Pipeline Systems
The seismic design guidelines recommended by ASCE (Nyman, et al, 1984) for oil
and gas pipeline systems are in many ways similar to the principles used in the design for
other important facilities. For important pipeline systems, the design should be based on
two-level earthquake hazard:
• The Probable Design Earthquake (PDE), the lower level, is generally associated with a
return period of 50 to 100 years.
• The Contingency Design Earthquake (CDE), the higher level, is represented by an
event with a return period of about 200 to 500 years. The general performance
requirements of the pipeline facilities under the two design events are also similar to
those for other facilities.
18

19
2.4 Proposed Seismic Design Philosophy for Tunnel
Structures
Two-Level Design Criteria
Based on the discussion presented above, it is apparent that current seismic design
philosophy for many civil engineering facilities has advanced to a state that dual (two-level)
design criteria are required. Generally speaking, the higher design level is aimed at
life safety while the lower level is intended for continued operation (i.e., an economical
design goal based on risk considerations). The lower-level design may prove to be a
good investment for the lifetime of the structures.
The two-level design criteria approach is recommended to ensure that transportation
tunnels constructed in moderate to high seismic areas represent functional adequacy and
economy while reducing life-threatening failure. This design philosophy has been
employed successfully in many of PB’s recent transportation tunnel projects (LA Metro,
Taipei Metro, Seattle Metro, and Boston Central Artery/Third Harbor Tunnel). In these
projects the two design events are termed as:
• The Operating Design Earthquake (ODE), defined as the earthquake event that can
reasonably be expected to occur during the design life of the facility (e.g., at least
once). The ODE design goal is that the overall system shall continue operating during
and after an ODE and experience little to no damage.
• The Maximum Design Earthquake (MDE), defined as an event that has a small
probability of exceedance during the facility life (e.g., 5 percent). The MDE design
goal is that public safety shall be maintained during and after an MDE.
Note, however, that the design criteria aimed at saving lives alone during a
catastrophic earthquake are sometimes considered unacceptable. There are cases
where more stringent criteria are called for under the maximum design earthquake, such
as requiring rapid repairs with relatively low cost. A good example would be the existing
San Francisco BART structures. As described in Chapter 1, BART warrants such stringent
criteria because it has an incalculable value as possibly the only reliable direct public
transportation system in the aftermath of a catastrophic earthquake.
Therefore, the actual acceptable risk and the performance goals during and after an
MDE depend on the nature and the importance of the facility, public safety and social
concerns, and potential direct and indirect losses.

Loading Criteria
Maximum Design Earthquake (MDE). Given the performance goals of the MDE (i.e.,
public safety), the recommended seismic loading combinations using the load factor
design method are as follows:
For Cut-and-Cover Tunnel Structures
(Eq. 2-1)
U = D + L + E1+ E2 +EQ
Where U = required structural strength capacity
D = effects due to dead loads of structural components
L = effects due to live loads
E1 = effects due to vertical loads of earth and water
E2 = effects due to horizontal loads of earth and water
EQ = effects due to design earthquake (MDE)
For Mined (Circular) Tunnel Lining
(Eq. 2-2)
U = D + L + EX +H + EQ
where U, D, L, and EQ are as defined in Equation 2-1
EX = effects of static loads due to excavation (e.g., O’Rourke, 1984)
H = effects due to hydrostatic water pressure
Comments on Loading Combinations for MDE
• The structure should first be designed with adequate strength capacity under static
loading conditions.
• The structure should then be checked in terms of ductility as well as strength when
earthquake effects, EQ, are considered. The “EQ” term for conventional surface
structure design reflects primarily the inertial effect on the structures. For tunnel
structures, the earthquake effect is governed by the displacements/deformations
imposed on the tunnels by the ground.
• In checking the strength capacity, the effects of earthquake loading should be
20

expressed in terms of internal moments and forces, which can be calculated
according to the lining deformations (distortions) imposed by the surrounding ground.
If the “strength” criteria expressed by Equation 2-1 or 2-2 can be satisfied based on
elastic structural analysis, no further provisions under the MDE are required.
Generally the strength criteria can easily be met when the earthquake loading intensity
is low (i.e., in low seismic risk areas) and/or the ground is very stiff.
• If the flexural strength of the tunnel lining, using elastic analysis and Equation 2-1 or 2-
2, is found to be exceeded (e.g., at certain joints of a cut-and-cover tunnel frame), one
of the following two design procedures should be followed:
(1) Provide sufficient ductility (using proper detailing procedure) at the critical
locations of the lining to accommodate the deformations imposed by the ground
in addition to those caused by other loading effects (see Equations 2-1 and 2-2).
The intent is to ensure that the structural strength does not degrade as a result of
inelastic deformations and the damage can be controlled at an acceptable level.
In general the more ductility is provided, the more reduction in earthquake forces
(the “EQ” term) can be made in evaluating the required strength, U. As a rule of
thumb, the force reduction factor can be assumed equal to the ductility provided.
This reduction factor is similar by definition to the response modification factor
used in bridge design code (AASHTO).
Note, however, that since an inelastic “shear” deformation may result in strength
degradation, it should always be prevented by providing sufficient shear
strengths in structure members, particularly in the cut-and-cover rectangular
frame.
(2) Re-analyze the structure response by assuming the formation of plastic hinges at
the joints that are strained into inelastic action. Based on the plastic-hinge
analysis, a redistribution of moments and internal forces will result.
If new plastic hinges are developed based on the results, the analysis is re-run by
incorporating the new hinges (i.e., an iterative procedure) until all potential plastic
hinges are properly accounted for. Proper detailing at the hinges is then carried
out to provide adequate ductility. The structural design in terms of required
strength (Equations 2-1 and 2-2) can then be based on the results from the
plastic-hinge analysis.
As discussed earlier, the overall stability of tunnel structures during and after the
MDE has to be maintained. Realizing that the structures also must have sufficient
capacity (besides the earthquake effect) to carry static loads (e.g., D, L, E1, E2
and H terms), the potential modes of instability due to the development of plastic
21

hinges (or regions of inelastic deformation) should be identified and prevented
(Monsees, 1991; see Figure 21 for example).
• The strength reduction factor, f, used in the conventional design practice may be too
conservative, due to the inherently more stable nature of underground structures
(compared to surface structures), and the transient nature of the earthquake loading.
• For cut-and-cover tunnel structures, the evaluation of capacity using Equation 2-1
should consider the uncertainties associated with the loads E1 and E2, and their worst
combination. For mined circular tunnels (Equation 2-2), similar consideration should
be given to the loads EX and H.
• In many cases, the absence of live load, L, may present a more critical condition than
when a full live load is considered. Therefore, a live load equal to zero should also be
used in checking the structural strength capacity using Equations 2-1 and 2-2.
Operating Design Earthquake (ODE). For the ODE, the seismic design loading
combination depends on the performance requirements of the structural members.
Generally speaking, if the members are to experience little to no damage during the lower-level
event (ODE), the inelastic deformations in the structure members should be kept low.
The following loading criteria, based on load factor design, are recommended:
For Cut-and-Cover Tunnel Structures
(Eq. 2-3)
where D, L, E1, E2, EQ, and U are as defined in Equation 2-1.
b1 = 1.05 if extreme loads are assumed for E1 and E2 with little uncertainty.
Otherwise, use b1 = 1.3.
For Mined (Circular) Tunnel Lining
(Eq. 2-4)
U =1.05D +1.3L +b2 EX +H ÊË
where D, L, EX, H, EQ, and U are as defined in Equation 2-2.
b2 = 1.05 if extreme loads are assumed for E1 and E2 with little uncertainty.
Otherwise, use b2 = 1.3.
ˆ¯
+1.3EQ
U =1.05D +1.3L +b1 E1 +E2 ÊË
ˆ¯
+1.3EQ
22

Comments on Loading Combinations for ODE
• The structure should first be designed with adequate strength capacity under static
loading conditions.
• For cut-and-cover tunnel structures, the evaluation of capacity using Equation 2-3
should consider the uncertainties associated with the loads E1 and E2, and their worst
combination. For mined circular tunnels (Equation 2-4), similar consideration should
be given to the loads EX and H.
When the extreme loads are used for design, a smaller load factor is recommended to
avoid unnecessary conservatism. Note that an extreme load may be a maximum load
or a minimum load, depending on the most critical case of the loading combinations.
Use Equation 2-4 as an example. For a deep circular tunnel lining, it is very likely that
the most critical loading condition occurs when the maximum excavation loading, EX,
is combined with the minimum hydrostatic water pressure, H. For a cut-and-cover
tunnel, the most critical seismic condition may often be found when the maximum
lateral earth pressure, E2, is combined with the minimum vertical earth load, E1. If a
very conservative lateral earth pressure coefficient is assumed in calculating the E2,
the smaller load factor b1 = 1.05 should be used.
• Redistribution of moments (e.g., ACI 318) for cut-and-cover concrete frames is
recommended to achieve a more efficient design.
• If the “strength” criteria expressed by Equation 2-3 or 2-4 can be satisfied based on
elastic structural analysis, no further provisions under the ODE are required.
• If the flexural strength of the tunnel lining, using elastic analysis and Equation 2-3 or 2-
4, is found to be exceeded, the structure should be checked for its ductility to ensure
that the resulting inelastic deformations, if any, are small. If necessary, the structure
should be redesigned to ensure the intended performance goals during the ODE.
• Zero live load condition (i.e., L = 0) should also be evaluated in Equations 2-3 and 2-4.
23

3.0 RUNNING LINE TUNNEL DESIGN
25

27
3.0 RUNNING LINE TUNNEL DESIGN
3.1 Overview
Discussions of the earthquake shaking effect on underground tunnels, specifically the
“EQ” term in Equations 2-1 through 2-4, are presented in a quantitative manner in this
chapter and in Chapters 4 and 5.
The response of tunnels to seismic shaking motions may be demonstrated in terms of
three principal types of deformations (Owen and Scholl, 1981):
• Axial
• Curvature
• Ovaling (for circular tunnels) or racking (for rectangular tunnels such as cut-and-cover
tunnels)
The first two types — axial and curvature — are considered in this chapter. Analytical
work developed in previous studies for tunnel lining design is presented. The work is
applicable to both circular mined tunnels and rectangular cut-and-cover tunnels.
Discussions of the third type — the ovaling effect on circular tunnels and the racking
effect on rectangular tunnels — are presented in detail in Chapters 4 and 5, respectively.
3.2 Types of Deformations
Axial and Curvature Deformations
Axial and curvature deformations develop in a horizontal or nearly horizontal linear
tunnel (such as most tunnels) when seismic waves propagate either parallel or obliquely to
the tunnel. The tunnel lining design considerations for these types of deformations are
basically in the longitudinal direction along the tunnel axis.
Figure 3 shows the idealized representations of axial and curvature deformations. The
general behavior of the linear tunnel is similar to that of an elastic beam subject to
deformations or strains imposed by the surrounding ground.

Figure 3.
Axial and Curvature Deformations
(Source: Owen and Scholl, 1981)
28

Ovaling or Racking Deformations
The ovaling or racking deformations of a tunnel structure may develop when waves
propagate in a direction perpendicular or nearly perpendicular to the tunnel axis, resulting
in a distortion of the cross-sectional shape of the tunnel lining. Design considerations for
this type of deformation are in the transverse direction.
Figure 4 shows the ovaling distortion and racking deformation associated with circular
tunnels and rectangular tunnels, respectively. The general behavior of the lining may be
simulated as a buried structure subject to ground deformations under a two-dimensional,
plane-strain condition.
Ovaling and racking deformations may be caused by vertically, horizontally or
obliquely propagating seismic waves of any type. Many previous studies have suggested,
however, that the vertically propagating shear wave is the predominant form of earthquake
loading that governs the tunnel lining design against ovaling/racking. The following
reasons are given:
• Ground motion in the vertical direction is generally considered less severe than its
horizontal component. Typically, vertical ground motion parameters are assumed to
be 1/2 to 2/3 of the horizontal ones. (Note that a vertically propagating shear wave
causes the ground to shake in the horizontal direction.) This relation is based on
observation of California earthquakes, which are most commonly of the strike-slip
variety in which horizontal motion predominates.
For thrust faults, in which one rock block overrides another, vertical effects may equal
or exceed the horizontal ones. The effects of thrust faulting are usually more
localized, however, than those of the strike-slip faulting, and they are attenuated more
rapidly with distance from the focus.
• For tunnels embedded in soils or weak media, the horizontal motion associated with
vertically propagating shear waves tends to be amplified. In contrast, the ground
strains due to horizontally propagating waves are found to be strongly influenced by
the ground strains in the rock beneath. Generally, the resulting strains are smaller
than those calculated using the properties of the soils.
29

Figure 4.
Ovaling and Racking Deformations
30

3.3 Free-Field Axial and Curvature Deformations
Background
The intensity of earthquake ground motion is described by several important
parameters, including peak acceleration, peak velocity, peak displacement, response
spectra, duration and others. For aboveground structures, the most widely used measure
is the peak ground acceleration and the design response spectra, as the inertial forces of
the structures caused by ground shaking provide a good representation of earthquake
loads.
Peak ground acceleration is not necessarily a good parameter, however, for
earthquake design of underground structures such as tunnels, because tunnel structures
are more sensitive to the distortions of the surrounding ground than to the inertial effects.
Such ground distortions — referred to in this report as free-field deformations/strains —
are the ground deformations/strains caused by the traveling seismic waves without the
structures being present. The procedure used to derive these deformations/strains is
discussed below.
A Practical Approach to Describing Ground Behavior
To describe the free-field ground behavior rigorously, even without the consideration
of ground structure interaction, is an extremely complex problem that would generally
require a three-dimensional dynamic analysis for solution. The earthquake source
characteristics and the transmission paths of various types of waves should also be
included in the model. This type of complex analysis, however, is rarely justified
economically.
For practical purposes, a simplified approach was proposed by Newmark (1968) and
has been considered by others (Sakurai and Takahashi, 1969; Yeh, 1974; and Agrawal et.
al, 1983). This approach is based on theory of wave propagation in homogeneous,
isotropic, elastic media. The ground strains are calculated by assuming a harmonic wave
of any wave type propagating at an angle (angle of incidence) with respect to the axis of a
planned structure.
Figure 5 (Kuesel, 1969) represents free-field ground deformations along a tunnel axis
due to a sinusoidal shear wave with a wavelength, L, a displacement amplitude, D, and an
angle of incidence, q. A conservative assumption of using the most critical angle of
incidence, and therefore the maximum values of strain, is often made, because the angle
of incidence for the predominant earthquake waves cannot be determined reliably.
31

Axial
Displacement
of Soil
Figure 5.
Geometry of a Sinusoidal Shear Wave Oblique to Axis of Tunnel
(Source: SFBARTD, 1960)
32
Axis of Tunnel
Transverse
Displacement
of Soil

33
Simplified Equations for Axial Strains and Curvature
Using the simplified approach, the free-field axial strains and curvature due to shear
waves and Rayleigh waves (surface waves) can be expressed as a function of angle of
incidence, as shown in Table 1. The most critical angle of incidence and the maximum
values of the strains are also included in the table.
Equations caused by compressional P-waves are also available, but it is generally
considered that they would not control the design. It is difficult to determine which type of
wave will dominate due to the complex nature of the characteristics associated with
different wave types. Generally, strains produced by Rayleigh waves may govern only
when the site is at a large distance from the earthquake source and the structure is built at
shallow depth.
Application of the strain equations presented in Table 1 requires knowledge of:
• The effective wave propagation velocity
• The peak ground particle velocity
• The peak ground particle acceleration
The peak velocity and acceleration can be established through empirical methods,
field measurements, or site-specific seismic exposure studies. The effective wave
propagation velocity in rock can be determined with reasonable confidence from in-situ
and laboratory tests.
Estimating the effective wave propagation velocity in soil overburden presents the
major difficulty. Previous studies have shown that, except possibly for vertically
propagating shear waves, the use of soil properties in deriving the wave velocity in soil
overburden may be overly conservative.
It has been suggested that for horizontally or obliquely propagating waves the
propagation velocities in soil overburden are affected significantly by the velocities in the
underlying rock. That is to say, the actual velocity values in the soils may be much higher
than those calculated based on the soil properties alone (Hadjian and Hadley, 1981). This
phenomenon is attributable to the problem of deformation compatibility. The motion of a
soil particle due to a horizontally propagating wave above the rock cannot differ greatly
from the motion of the rock, unless the soil slides on top of the rock (a very unlikely
occurrence) or the soil liquifies. For a very deep (thick) soil stratum, however, the top of
the soil stratum is less coupled to the rock and is more free to follow a motion that is
determined by its own physical properties.

q = Angle of Incidence with respect to Tunnel Axis
r = Radius of Curvature
VS, VR = Peak Particle Velocity for Shear Wave and Rayleigh Wave, respectively
CS, CR = Effective Propagation Velocity for Shear Wave and Rayleigh Wave,
respectively
AS, AR = Peak Particle Acceleration for Shear Wave and Rayleigh Wave, respectively
34
Wave Type Longitudinal Strain
(Axial)
Curvature
Shear
General
Form
Wave
Maximum
Value
Rayleigh
General
Form
Wave
Maximum
Value
e =
Vs
Cs
sinqcosq
1
r
ÊË
ˆ¯
=
As
Cs
2 cos3 q
emax =
Vs
2Cs
for q = 45
1
r
ÊË
ˆ ¯max =
As
Cs
2 for q = 0
e =
VR
CR
cos2q
1
r
ÊË
ˆ¯
=
AR
CR
2 cos2 q
emax =
VR
CR
for q = 0
1
r
ÊË
ˆ ¯max =
AR
CR
2 for q = 0
Table 1.
Free-Field Ground Strains

3.4 Design Conforming to Free-Field Axial and Curvature
Deformations
Background and Assumptions
The free-field ground strain equations, originally developed by Newmark (Table 1),
have been widely used in the seismic design of underground pipelines. This method has
also been used successfully for seismic design of long, linear tunnel structures in several
major transportation projects (Monsees, 1991; Kuesel, 1969).
When these equations are used, it is assumed that the structures experience the
same strains as the ground in the free-field. The presence of the structures and the
disturbance due to the excavation are ignored. This simplified approach usually provides
an upper-bound estimate of the strains that may be induced in the structures by the
traveling waves. The greatest advantage of this approach is that it requires the least
amount of input.
Underground pipelines, for which this method of analysis was originally developed,
are flexible because of their small diameters (i.e., small bending rigidity), making the free-field
deformation method a simple and reasonable design tool. For large underground
structures such as tunnels, the importance of structure stiffness sometimes cannot be
overlooked. Some field data indicated that stiff tunnels in soft soils rarely experience
strains that are equal to the soil strains (Nakamura, Katayama, and Kubo, 1981). A
method to consider tunnel stiffness will be presented and discussed later in Section 3.5.
Design Example 1: The Los Angeles Metro
For the purpose of illustration, a design example modified from the seismic design
criteria for the LA Metro project (SCRTD, 1984) is presented here. In this project, it was
determined that a shear wave propagating at 45 degree (angle of incidence) to the tunnel
axis would create the most critical axial strain within the tunnel structure. Although a P-wave
(compressional wave) traveling along the tunnel axis might also produce a similar
effect, it was not considered because:
• Measurement of P-wave velocity can be highly misleading, particularly when a soil
deposit is saturated with water (Monsees, 1991).
• The magnitudes of soil strains produced by a nearly horizontally propagating P-wave
are generally small and about the same as those produced in the underlying rock
and, therefore, not as critical as the shear-wave generated axial strains (SFBART,
1960). This phenomenon was discussed previously in Section 3.3.
35

36
Other assumptions and parameters used in this example are:
• Design Earthquake Parameters: Peak Ground Acceleration,
As = 0.6 (Maximum Design Earthquake, MDE)
• Peak Ground Velocity, Vs = 3.2 ft/sec
• Soil surrounding Tunnel: Fernando Formation
• Effective Shear Wave Velocity: Cs = 1360 ft/sec (in Fernando Formation under MDE)
• Tunnel Structure: Cast-in-place circular segmented reinforced lining,
with Radius R =10 feet
From Table 1, the combined maximum axial strain and curvature strain would be:
As the results of calculations indicate, the curvature (bending) component (0.000037)
is, in general, relatively insignificant for tunnel structures under seismic loads. According
to the LA Metro criteria, the maximum usable compression strain (under MDE) in the
concrete lining is eallow =0.002, since the strain is almost purely axial. With emax < eallow,
the lining is considered adequate in compression under the Maximum Design Earthquake
(MDE).
The calculated maximum axial strain (=0.00122) is cyclic in nature. When tension is in
question, a plain concrete lining would obviously crack. The assumed lining is reinforced,
however, and the opening of these cracks is transient due to the cyclic nature of seismic
waves. As long as no permanent ground deformation results, these cracks will be closed
by the reinforcing steel at the end of the shaking. Even in the unreinforced concrete lining
cases, the lining generally is considered adequate as long as:
• The crack openings are small and uniformly distributed
• The resulting tension cracks do not adversely affect the intended performance goals
of the lining
emax = ±
Vs
2Cs
±
AsR
Cs
2 cos3q
= ±
3.2
2x1360
±
0.6x32.2x10
(1360)2 cos345
= ± 0.00118 ± 0.000037
= ± 0.00122

37
Applicability of the Free-Field Deformation Approach
The example presented above demonstrates the simplicity of the free-field
deformation approach. Because it is an upper-bound assessment of the tunnel response,
it often becomes the first tool an engineer would use to verify the adequacy of his design.
This approach offers a method for verification of a design rather than a design itself.
Note, however, that this method is:
• Pertinent to a tunnel structure that is flexible relative to its surrounding medium, such
as all tunnels in rock and most tunnels in stiff soils. In this case it is reasonable to
assume that the tunnel deforms according to its surrounding medium.
• Not desirable for situations involving stiff structures buried in soft soil, because under
this condition, the calculated ground deformations may be too great (due to the soft
nature of the soil) for the stiff structures to realistically accommodate. Once the
calculated ground strain exceeds the allowable strain of the lining material, there is
very little an engineer can do to improve his design.
For instance, if the effective shear wave velocity of the previous example is reduced to
350 ft/sec to reflect a much softer soil deposit, the tunnel lining will then be subjected to a
combined maximum axial strain of 0.0052 in compression (see Design Example 2 in the
next section). It is essentially unrealistic to provide an adequate concrete lining design
resisting an axial strain of this amount. If the free-field deformation approach were used in
this case, it appears that the only solution to this problem would be to provide needless
flexible joints, forming a chainlink-like tunnel structure to accommodate the ground
deformation.
In the next section, a design approach considering the tunnel-ground interaction
effect is presented. This design approach, based on results from previous studies, may
effectively alleviate the design difficulty discussed above.
3.5 Tunnel-Ground Interaction
When it is stiff in its longitudinal direction relative to its surrounding soils, the tunnel
structure resists, rather than conforms to, the deformations imposed by the ground.
Analysis of tunnel-ground interaction that considers both the tunnel stiffness and ground
stiffness plays a key role in finding the tunnel response. With the computation capability of
today’s computers, this problem may be solved numerically using sophisticated computer
codes.
For practical purposes, however, a simplified procedure is desirable and has been
sought in previous studies (SFBARTD, 1960; Kuribayashi, et al, 1974; and St. John, et al,

38
1987). In general, the tunnel-ground system is simulated as an elastic beam on an elastic
foundation, with the theory of wave propagating in an infinite, homogeneous, isotropic
medium. When subjected to the axial and curvature deformations caused by the traveling
waves in the ground, the tunnel will experience the following sectional forces (see Figure 6):
• Axial forces, Q, on the cross-section due to the axial deformation
• Bending moments, M, and shear forces, V, on the cross-section due to the curvature
deformation
Simplified Interaction Equations
Maximum Axial Force: Qmax. Through theoretical derivations, the resulting maximum
sectional axial forces caused by a shear wave with 45 degree angle of incidence can be
obtained:
(Eq. 3-1)
Where L = wavelength of an ideal sinusoidal shear wave
2 D
Ka =longitudinal spring coefficient of medium (in force per unit deformation per
unit length of tunnel)
D = free-field displacement response amplitude of an ideal sinusoidal shear
wave
Ec = modulus of elasticity of tunnel lining
Ac = cross-section area of tunnel lining
The calculated maximum axial force, Qmax, shall not exceed an upper limit defined by
the ultimate soil drag resistance in the longitudinal direction. This upper limit is expressed
as:
(Eq. 3-2)
Qlimit =
where f = ultimate friction force (per unit length of tunnel) between the tunnel and
the surrounding medium
fL
4
Qmax =
KaL
2p
1 +2
Ka
EcAc
Ê
Ë
ˆ¯
L
2p
Ê
Ë
ˆ¯

39
Figure 6.
Sectional Forces Due to Curvature and Axial Deformations
(Source: Owen and Scholl, 1981)

Maximum Bending Moment, Mmax. The bending moment resulting from curvature
deformations is maximized when a shear wave is traveling parallel to the tunnel axis (i.e.,
with an angle of incidence equal to zero). The mathematical expression of the maximum
bending moment is:
40
(Eq. 3-3)
Mmax =
Kt
L
2p
Ê
ÁÁ
Ë
2
ˆ
˜˜
¯
1+
Kt
EcIc
Ê
Ë
Ê
Ë
ˆ¯
L
2p
where L, Ec and D are as defined in Equation 3-1
4 D
ˆ¯
Ic = moment of inertia of the tunnel section
Kt = transverse spring coefficient of medium (in force per unit deformation
per unit length of tunnel).
Maximum Shear Force, Vmax. The maximum shear force corresponding to the maximum
bending moment is derived as:
(Eq. 3-4)
V max =
KtL
2p
1+
Kt
EcIc
Ê
Ë
Ê
Ë
ˆ¯
L
2p
4 D = Mmax
ˆ¯
where L, Ec, Ic, Kt and D are as defined in Equation 3-3.
2p
L
Comments on the Interaction Equations
• The tunnel-ground interaction effect is explicitly accounted for in these formulations.
The ground stiffness and the tunnel stiffness are represented by spring coefficients
(Ka or Kt) and sectional modulus (EcAc or EcIc), respectively.
• The application of these equations is necessary only when tunnel structures are built
in soft ground. For structures in rock or stiff soils, the evaluation based on the free-field
ground deformation approach presented in Section 3.3 will, in general, be
satisfactory.
• Equations 3-1, 3-3 and 3-4 are general mathematical forms. Other expressions of the
maximum sectional forces exist in the literature. The differences are primarily due to
the further maximization of the sectional forces with respect to the wavelength, L. For
instance:

41
- In the JSCE (Japanese Society of Civil Engineers) Specifications for Earthquake
Resistant Design of Submerged Tunnels, the values of wavelength that will
maximize Equations 3-1, 3-3 and 3-4 are determined and substituted back into
each respective equation to yield the maximum sectional forces.
- St. John and Zahran (1987) suggested a maximization scheme that is similar to
the Japanese approach except that the spring coefficients (Ka or Kt) are assumed
to be functions of wavelength, L, in the maximization process.
Both of these approaches assume that the free-field ground displacement response
amplitude, D, is independent of the wavelength. This assumption sometimes may
lead to very conservative results, as the ground displacement response amplitude
generally decreases with the wavelength. It is, therefore, the author’s view that
Equations 3-1 through 3-4 presented in this section will provide a practical and
adequate assessment, provided that the values (or the ranges of the values) of L, D,
and Kt (or Ka) can be reasonably estimated.
A reasonable estimate of the wavelength can be obtained by
(Eq. 3-5)
L =T Cs
where T is the predominant natural period of the shear wave traveling in the soil
deposit in which the tunnel is built, and Cs is the shear wave propagation velocity
within the soil deposit.
Often, T can also be represented by the natural period of the site. Dobry, Oweis and
Urzua (1976) presented some procedures for estimating the natural period of a linear
or equivalent linear model of a soil site.
• The ground displacement response amplitude, D, should be derived based on site-specific
subsurface conditions by earthquake engineers. The displacement
amplitude represents the spatial variations of ground motions along a horizontal
alignment. Generally, the displacement amplitude increases as the wavelength, L,
increases. For example, the displacement spectrum chart prepared by Housner
(SFBARTD, 1960) for the SF BART project was expressed by D = 4.9 x 10-6 L1.4,
where the units of D and L are in feet. This spectrum is intended for tunnel tubes in
soft San Francisco Bay muds and was derived for a magnitude 8.2 earthquake on the
San Andreas fault. The equation shows clearly that:
- The displacement amplitude increases with the wavelength.
- For any reasonably given wavelength, the corresponding ground displacement

42
amplitude is relatively small. Using the given wavelength and the corresponding
displacement amplitude, the calculated free-field ground strains would be
significantly smaller than those calculated using the simplified equations shown in
Table 1. This suggests that it may be overly conservative to use the simplified
equations to estimate the axial and curvature strains caused by seismic waves
travelling in soils for tunnel design.
• With regard to the derivations of spring coefficients Ka and Kt, there is no consensus
among design engineers. The derivations of these spring coefficients differ from those
for the conventional beam on elastic foundation problems in that:
-The spring coefficients should be representative of the dynamic modulus of the
ground under seismic loads.
-The derivations should consider the fact that loading felt by the surrounding soil
(medium) is alternately positive and negative due to the assumed sinusoidal
seismic wave.
Limited information on this problem is available in the literature (SFBARTD 1960, St.
John and Zahrah, 1987 and Owen and Scholl, 1981). For preliminary design, it
appears that the expressions suggested by St. John and Zahrah (1987) should serve
the purpose:
(Eq. 3-6)
Kt = Ka =
16pGm(1 -vm)
(3 -4vm)
d
L
where Gm = shear modulus of the medium (see Section 4.2 in Chapter 4)
nm = Poisson’s radio of the medium
d = diameter (or equivalent diameter) of the tunnel
L = wavelength
• A review of Equations 3-1, 3-3 and 3-4 reveals that increasing the stiffness of the
structure (i.e., EcAc and EcIc), although it may increase the strength capacity of the
structure, will not result in reduced forces. In fact, the structure may attract more
forces as a result. Therefore, the designer should realize that strengthening of an
overstressed section by increasing its sectional dimensions (e.g., lining thickness)
may not always provide an efficient solution for seismic design of tunnel structures.
Sometimes, a more flexible configuration with adequate reinforcements to provide
sufficient ductility is a more desirable measure.

43
Design Example 2: A Linear Tunnel in Soft Ground
In this example, a tunnel lined with a cast-in-place circular concrete lining (e.g., a
permanent second-pass support) is assumed to be built in a soft soil site. The
geotechnical, structural and earthquake parameters are listed as follows:
Geotechnical Parameters:
- Effective shear wave velocity, CS =350 ft/sec.
- Soil unit weight, gt =110 pcf =0.110 kcf.
- Soil Poisson’s ratio, nm =0.5 (saturated soft clay).
- Soil deposit thickness over rigid bedrock, H =100 ft.
Structural Parameters:
- Lining thickness, t =1 ft.
- Lining diameter, d =20 ft.
- Lining moment of inertia, Ic = 0.5 x 3148 = 1574 ft4
(one half of the full section moment of inertia to account for concrete cracking and
nonlinearity during the MDE).
- Lining cross section area, Ac =62.8 ft2.
- Concrete Young’s Modulus, Ec =3600 ksi =518400 ksf.
- Concrete yield strength, fc =4000 psi.
- Allowable concrete compression strain under combined axial and bending
compression, eallow = 0.003 (during the MDE)
Earthquake Parameters (for the MDE):
- Peak ground particle acceleration in soil, As =0.6 g.
- Peak ground particle velocity in soil, Vs =3.2 ft/sec.

44
First, try the simplified equation as used in Design Example 1. The combined
maximum axial strain and curvature strain is calculated as:
The calculated maximum compression strain exceeds the allowable compression
strain of concrete (i.e., emax > eallow = 0.003).
Now use the tunnel-ground interaction procedure.
1. Estimate the predominant natural period of the soil deposit (Dobry, et al, 1976).
2. Estimate the idealized wavelength (Equation 3-5):
L =TxCs = 4H
=400 ft
3. Estimate the shear modulus of soil:
Gm =rCs
4. Derive the equivalent spring coefficients of the soil (Equation 3-6):
K = K =
16pG m
(1-nm )
(3-4nm)
d
L
=
16 x418.5 (1 -0.5)
(3 -4 x0.5)
x
20
400
p
=526 kips/ft
a t
2 =
0.110kcf
32.2
x3502= 418.5ksf
T =
4H
Cs
=
4x100'
350
=1.14 sec.
emax =±
Vs
2Cs
±
AsR
Cs
2
cos3q= ±
.3.2
2x350
±
0.6 32.2x10
( 350) 2
cos3 45o
=±0.0046 ±0.0006 = ±0.0052
x

Vs
2Cs
=
2pD
L
fi D = Da =0.291ft
4p2D
L2 fi D = Db = 0.226 ft 2
45
5. Derive the ground displacement amplitude, D:
As discussed before, the ground displacement amplitude is generally a function of the
wavelength, L. A reasonable estimate of the displacement amplitude must consider the
site-specific subsurface conditions as well as the characteristics of the input ground
motion. In this design example, however, the ground displacement amplitudes are
calculated in such a manner that the ground strains as a result of these displacement
amplitudes are comparable to the ground strains used in the calculations based on the
simplified free-field equations. The purpose of this assumption is to allow a direct and
clear evaluation of the effect of tunnel-ground interaction. Thus, by assuming a sinusoidal
wave with a displacement amplitude D and a wavelength L, we can obtain:
For free-field axial strain:
For free-field bending curvature:
6. Calculate the maximum axial force (Equation 3-1) and the corresponding axial strain
of the tunnel lining:
Qmax =
KaL
2p
1 +2
Ka
EcAc
ÊË
ˆ¯
L
2p
ÊË
2 Da
ˆ¯
=
526x400
2p
ÊË
1+ 2
526
ˆ¯
518400x62.8
400
2p
ÊË
2 x 0.291
ˆ¯
=8619kips
eaxial =
Qmax
Ec Ac
=
8619
518400x62.8
=0.00026
As
Cs
cos345o =

7. Calculate the maximum bending moment (Equation 3-3) and the corresponding
46
bending strain of the tunnel lining:
2
4
2
Ê
Ë
4
8. Compare the combined axial and bending compression strains to the allowable:
9. Calculate the maximum shear force (Equation 3-4) due to the bending curvature:
V max = Mmax x 2p
L
= 41539
2p
400
= 652kips
x
emax = eaxial+ ebending
= 0.00026 +0.00051
= 0.00077< eallow = 0.003
Mmax =
Kt
L
2p
Ê
Ë
ˆ¯
1 +
Kt
EcIc
Ê
Ë
ˆ¯
L
2p
Ê
Ë
ˆ¯
Db
=
526
400
2p
Ê
Ë
ˆ¯
1 +
526
518400x1574
ˆ¯
400
2p
Ê
Ë
ˆ¯
x0.226
=41539 k - ft
ebending =
Mmax R
EcIc
=
41539x10
518400x1574
=0.00051

47
10. Calculate the allowable shear strength of concrete during the MDE:
fVc =0.85x2 fcAshear
where f = shear strength reduction factor (0.85)
fc = yield strength of concrete (4000 psi)
Ashear = effective shear area = Ac/2
Note: Use of f= 0.85 for earthquake design may be very conservative.
Ashear
fi c =0.85x2 4000 x
62.8
2
x
144
1000
= 486 kips
fV
11. Compare the induced maximum shear force with the allowable shear resistance:
Vmax =625 kips > fVc = 486 kips
Although calculations indicate that the induced maximum shear force exceeds the
available shear resistance provided by the plain concrete, this problem may not be of
major concern in actual design because:
• The nominal reinforcements generally required for other purposes may provide
additional shear resistance during earthquakes.
• The ground displacement amplitudes, D, used in this example are very conservative.
Generally the spatial variations of ground displacements along a horizontal axis are
much smaller than those used in this example, provided that there is no abrupt
change in subsurface profiles.

3.6 Special Considerations
48
Through the design examples 1 and 2 presented above, it was demonstrated that
under normal conditions the axial and curvature strains of the ground were not critical to
the design of horizontally aligned linear tunnels. Special attention is required, however, in
the following situations:
• Unstable ground, including ground that is susceptible to landslide and/or liquefaction
• Faulting, including tectonic uplift and subsidence
• Abrupt changes in structural stiffness or ground conditions
Unstable Ground
It is generally not feasible to design a tunnel lining of sufficient strength to resist large
permanent ground deformations resulting from an unstable ground. Therefore, the proper
design measures in dealing with this problem should consider the following:
• Ground stabilization (e.g., compaction, draining, reinforcement, grouting, and earth
retaining systems)
• Removal and replacement of problem soils
• Reroute or deeper burial
Faulting
With regard to fault displacements, the best solution is to avoid any potential crossing
of active faults. If this is not possible, the general design philosophy is to design a tunnel
structure to accept and accommodate these fault displacements. For example, in the
North Outfall Replacement Sewer (NORS, City of Los Angeles) project, the amount of fault
displacement associated with an M=6.5 design earthquake on the Newport-Inglewood
fault was estimated to be about 8 inches at the crossing. To accommodate this
displacement, a design scheme using an oversized excavation and a compressible
backpacking material was provided. The backpacking material was designed to
withstand the static loads, yet be crushable under faulting movements to protect the pipe.

It is believed that the only transportation tunnel in the U.S. designed and constructed
to take into consideration potential active fault displacements is the Berkeley Hills Tunnel,
part of the San Francisco BART system. This horse-shoe-shaped tunnel was driven
through the historically active creeping Hayward Fault with a one-foot oversized
excavation. The purpose of the over-excavation was to provide adequate clearance for
rail passage even when the tunnel was distorted by the creeping displacements. Thus
rails in this section could be realigned and train services could be resumed quickly
afterward.
The tunnel was lined with concrete encased ductile steel ribs on two-foot centers. The
concrete encased steel rib lining is particularly suitable for this design because it provides
sufficient ductility to accommodate the lining distortions with little strength degradation.
The two projects described above have several common design assumptions that
allowed the special design to be feasible both technically and economically:
• The locations of the faults at crossings can be identified with acceptable uncertainty,
limiting the lengths of the structures that require such special design.
• The design fault displacements are limited to be within one foot.
The cost associated with special design may become excessively high when
significant uncertainty exists in defining the activities and locations of the fault crossings,
or when the design fault displacements become large (e.g., five feet). Faced with these
situations, designers as well as owners should re-evaluate and determine the performance
goals of the structures based on a risk-cost balanced consideration, and design should be
carried out accordingly.
Abrupt Changes in Structural Stiffness or Ground Conditions
These conditions include, but are not limited to, the following:
• When a regular tunnel section is connected to a station end wall or a rigid, massive
structure such as a ventilation building
• At the junctions of tunnels
• When a tunnel traverses two distinct geological media with sharp contrast in stiffness
• When tunnels are locally restrained from movements by any means (i.e., “hard spots”)
49

Generally, the solutions to these interface problems are to provide either of the following:
• A movable joint, such as the one used at the connection between the Trans-Bay Tube
and the ventilation building (Warshaw, 1968)
• A rigid connection with adequate strength and ductility
At these critical interfaces, structures are subjected to potential differential movements
due to the difference in stiffness of two adjoining structures or geological media.
Estimates of these differential movements generally require a dynamic analysis taking into
account the soil-structure interaction effect (e.g., SFBARTD, 1991). There are cases
where, with some assumptions, a simple free-field site response analysis will suffice. The
calculated differential movements provide necessary data for further evaluations to
determine whether special seismic joints are needed.
Once the differential movements are given, there are some simple procedures that
may provide approximate solutions to this problem. For example, a linear tunnel entering
a large station may experience a transverse differential deflection between the junction
and the far field due to the large shear rigidity provided by the end wall of the station
structure. If a conventional design using a rigid connection at the interface is proposed,
additional bending and shearing stresses will develop near the interface. These stress
concentrations can be evaluated by assuming a semi-infinite beam supported on an
elastic foundation, with a fixed end at the connection. According to Yeh (1974) and
Hetenyi (1976), the induced moment, M(x), and shear, V(x), due to the differential
transverse deflection, d, can be estimated as:
(Eq. 3-7)
(Eq. 3-8)
M(x) =
Kt
2l2 de - l x (sinlx -coslx)
V(x) =
Kt
l
-lxcos
de lx
l =
Kt
4EcIc
Ê
Ë
where x = distance from the connection
Ic = moment of inertia of the tunnel cross section
Ec = Young’s modulus of the tunnel lining
Kt = transverse spring coefficient of ground (in force per unit deformation per
unit length of tunnel)
ˆ¯ 1
4
50

Based on Equations 3-7 and 3-8, the maximum bending moment and shear force
occur at x=0 (i.e., at the connection). If it is concluded that an adequate design cannot be
achieved by using the rigid connection scheme, then special seismic (movable) joints
should be considered.
51

4.0 OVALING EFFECT ON CIRCULAR TUNNELS
53

4.0 OVALING EFFECT ON CIRCULAR TUNNELS
The primary purpose of this chapter is to provide methods for quantifying the seismic
ovaling effect on circular tunnel linings. The conventionally used simplified free-field
deformation method, discussed first, ignores the soil-structure interaction effects.
Therefore its use, as demonstrated by two examples, is limited to certain conditions.
A refined method is then presented that is equally simple but capable of eliminating the
drawbacks associated with the free-field deformation method. This refined method — built
from a theory that is familiar to most mining/underground engineers — considers the soil-structure
interaction effects. Based on this method, a series of design charts are developed
to facilitate the design process. The results are further validated through numerical analyses.
4.1 Ovaling Effect
As defined in Chapter 3, ovaling of a circular tunnel lining is primarily caused by
seismic waves propagating in planes perpendicular to the tunnel axis (see Figure 2).
Usually, it is the vertically propagating shear waves that produce the most critical ovaling
distortion of the lining. The results are cycles of additional stress concentrations with
alternating compressive and tensile stresses in the tunnel lining. These dynamic stresses
are superimposed on the existing static state of stress in the lining. Several critical modes
may result (Owen and Scholl, 1981):
• Compressive dynamic stresses added to the compressive static stresses may exceed
the compressive capacity of the lining locally.
• Tensile dynamic stresses subtracted from the compressive static stresses reduce the
lining’s moment capacity, and sometimes the resulting stresses may be tensile.
4.2 Free-Field Shear Deformations
As discussed in Chapter 3, the shear distortion of ground caused by vertically
propagating shear waves is probably the most critical and predominant mode of seismic
motions in many cases. It causes a circular tunnel to oval and a rectangular underground
structure to rack (sideways motion), as shown in Figure 3. Analytical procedures by
numerical methods are often required to arrive at a reasonable estimate of the free-field
shear distortion, particularly for a soil site with variable stratigraphy. Many computer
codes with variable degree of sophistication are available (e.g., SHAKE, 1972; FLUSH,
1975; and LINOS, 1991).
55

The most widely used approach is to simplify the site geology into a horizontally
layered system and to derive a solution using one-dimensional wave propagation theory
(Schnabel, Lysmer, and Seed, 1972). The resulting free-field shear distortion of the
ground from this type of analysis can be expressed as a shear strain distribution or shear
deformation profile versus depth. An example of the resulting free-field shear distortion for
a soil site using the computer code SHAKE is presented in Figure 7.
Simplified Equation for Shear Deformations
For a deep tunnel located in relatively homogeneous soil or rock, the simplified
procedure by Newmark (presented in Table 1) may also provide a reasonable estimate.
Here, the maximum free-field shear strain, gmax, can be expressed as:
(Eq. 4-1)
gmax =
where Vs = peak particle velocity
Vs
Cs
Cs = effective shear wave propagation velocity
The values of Cs can be estimated from in-situ and laboratory tests. An equation
relating the effective propagation velocity of shear waves to effective shear modulus, Gm,
is expressed as:
(Eq. 4-2)
C =
Gm
r
s
where r = mass density of the ground
It is worth noting that both the simplified procedure and the more refined SHAKE
analysis require the parameters Cs or Gm as input. The propagation velocity and the shear
modulus to be used should be compatible with the level of shear strains that may develop
in the ground under design earthquake loading. This is particularly critical for soil sites
due to the highly non-linear behavior of soils. The following data are available:
• Seed and Idriss (1970) provide an often used set of laboratory data for soils giving the
effective shear wave velocity and effective shear modulus as a function of shear
strain.
• Grant and Brown (1981) further supplemented the data sets with results from a series
of field geophysical measurements and laboratory testing conducted for six soil sites.
56

57
Figure 7.
Free-Field Shear Distortions of Ground Under Vertically
Propagating Shear Waves

4.3 Lining Conforming to Free-Field Shear Deformations
When a circular lining is assumed to oval in accordance with the deformations
imposed by the surrounding ground (e.g., shear), the lining’s transverse sectional stiffness
is completely ignored. This assumption is probably reasonable for most circular tunnels in
rock and in stiff soils, because the lining stiffness against distortion is low compared with
that of the surrounding medium. Depending on the definition of “ground deformation of
surrounding medium,” however, a design based on this assumption may be overly
conservative in some cases and non-conservative in others. This will be discussed further
as follows.
Shear distortion of the surrounding ground, for this discussion, can be defined in two
ways. If the non-perforated ground in the free-field is used to derive the shear distortion
surrounding the tunnel lining, the lining is to be designed to conform to the maximum
diameter change, DD , shown in Figure 8. The diametric strain of the lining for this case
can be derived as:
(Eq. 4-3)
DD
D
= ±
where D = the diameter of the tunnel
g max
2
gmax = the maximum free-field shear strain
On the other hand, if the ground deformation is derived by assuming the presence of
a cavity due to tunnel excavation (Figure 9, for perforated ground), then the lining is to be
designed according to the diametric strain expressed as:
(Eq. 4-4)
DD
D
= ± 2gmax (1 -vm)
where nm = the Poisson’s Ratio of the medium
Equations 4-3 and 4-4 both assume the absence of the lining. In other words, tunnel-ground
interaction is ignored.
Comparison between Equations 4-3 and 4-4 shows that the perforated ground
deformation would yield a much greater distortion than the non-perforated, free-field
ground deformation. For a typical ground medium, an engineer may encounter solutions
provided by Equations 4-3 and 4-4 that differ by a ratio ranging from 2 to about 3. By
intuition:
58

59
Figure 8.
Free-Field Shear Distortion of Ground
(Non-Perforated Medium)
Figure 9.
Shear Distortion of Perforated Ground
(Cavity in-Place)

• Equation 4-4, the perforated ground deformation, should serve well for a lining that
has little stiffness (against distortion) in comparison to that of the medium.
• Equation 4-3, on the other hand, should provide a reasonable distortion criterion for a
lining with a distortion stiffness equal to the surrounding medium.
It is logical to speculate further that a lining with a greater distortion stiffness than the
surrounding medium should experience a lining distortion even less than that calculated
by Equation 4-3. This latest case may occur when a tunnel is built in soft to very soft soils.
The questions that may be raised are:
• How important is the lining stiffness as it pertains to the engineering design?
• How should the lining stiffness be quantified relative to the ground?
• What solutions should an engineer use when the lining and ground conditions differ
from those where Equations 4-3 and 4-4 are applicable?
In the following sections (4.4 and 4.5), answers to these questions are presented.
4.4 Importance of Lining Stiffness
Compressibility and Flexibility Ratios
To quantify the relative stiffness between a circular lining and the medium, two ratios
designated as the compressibility ratio, C, and the flexibility ratio, F (Hoeg, 1968, and Peck
et al., 1972) are defined by the following equations:
(Eq. 4-5)
(Eq. 4-6)
Compressibility Ratio, C =
Flexibility Ratio, F =
Em (1 -v1
E1t (1+vm) (1- 2vm)
Em (1-v1
2) R
2) R3
6E1I (1+ vm)
where Em = modulus of elasticity of the medium
nm = Poisson’s Ratio of the medium
El = the modulus of elasticity of the tunnel lining
nl = Poisson’s Ratio of the tunnel lining
60

R = radius of the tunnel lining
t = thickness of the tunnel lining
I = moment of inertia of the tunnel lining (per unit width)
Of these two ratios, it is often suggested that the flexibility ratio is the more important
because it is related to the ability of the lining to resist distortion imposed by the ground.
As will be discussed later in this chapter, the compressibility ratio also has an effect on the
lining thrust response.
The following examples on the seismic design for several tunnel-ground
configurations are presented to investigate the adequacy of the simplified design
approach presented in the previous section.
Example 1
The first illustrative example is a tunnel cross-section from the LA Metro project. The
ground involved is an old alluvium deposit with an effective shear wave propagation
velocity, Cs, equal to 1000 ft/sec. The peak shear wave particle velocity, Vs, according to
the design criteria, is 3.4 ft/sec.
Using Equation 4-1, the maximum free-field shear strain, gmax , is calculated to be
0.0034. The reinforced cast-in-place concrete lining properties and the soil properties are
assumed and listed in the following table.
Lining Properties Soil Properties
R = 9.5 feet Em = 7200 ksf
t = 8.0 inches nm = 0.333
El/(1- nl
2) = 662400 ksf
I = 0.0247 ft4/ft
Flexibility Ratio, F = 47
Compressibility Ratio, C = 0.35
Note that uncertainties exist in the estimates of many of the geological and structural
parameters. For instance:
• The effective shear wave propagation velocity in the old alluvium may have an
uncertainty of at least 20 percent.
• Uncertainty up to 40 percent may also be applied to the estimates of Em.
61

• The moment of inertia, I, for a cracked lining section, or for a segmental lining with
staggered joints in successive rings, may be considerably less than that for the typical
cross section of a segment as used in this calculation example. (See Section 4.5 for a
means of estimating the effective moment of inertia, Ie.)
It would be desirable, therefore, to define the ranges of the values considering these
uncertainties in the actual design cases.
The LA Metro project has adopted Equation 4-4 as the criterion for ovaling of the
running lines (SCRTD, 1984). Therefore, a maximum diametric strain, DD/D, of 0.00453 is
obtained. The maximum combined bending strain and thrust compression strain as a
result of this diametric strain is calculated, with some simple assumptions based on ring
theory, by using the following formulation:
(Eq. 4-7)
etotal =
Vs
Cs
Ê
Ë
ÏÌÓ
3(1-vm)
t
R
ˆ¯
+
1
2
R
t
Ê
Ë
È
ˆ¯
Em(1-v1
2)
E1(1+vm)
Î Í
¸˝˛
˘
° ˙
= 0.00061
To verify the accuracy of the results, a numerical analysis using finite difference code
is performed. No-slip interface between the lining and the surrounding ground is assumed
in the analysis. A more detailed description of this modeling is presented in Section 4.5.
Results from the finite difference analysis yield:
• A maximum diametric strain of 0.0038
• A combined maximum total compression strain in the lining of about 0.0006
The excellent agreement between the simplified approach using Equation 4-4 and the
refined numerical analysis is explained by the flexibility ratio (F=47) of the ground-lining
system. A flexibility ratio of this magnitude suggests that the lining should be flexible even
when compared to ground with a cavity in it, and therefore should conform to the
perforated ground deformation.
Example 2
In this example, the tunnel is assumed to be built in a very soft soil deposit. The
cross-sectional properties of the lining and the surrounding ground are shown in the
following table. Note that these properties are made in order to result in a flexibility ratio
equal to 1.0.
62

Lining Properties Soil Properties
R = 10 feet Em = 325 ksf
t = 12 inches nm = 0.25
El/(1-nl
2) = 518400 ksf
I = 0.0833 ft4/ft
Flexibility Ratio, F = 1.0
Compressibility Ratio, C = 0.01
It is further assumed that the free-field maximum shear strain, gmax = 0.008, is
obtained from one-dimensional site response analysis using SHAKE program. If Equation
4-4 is used, the maximum diametric strain, DD/D , of the lining is calculated to be 0.012.
With this diameter change, the lining will be subject to a maximum bending strain of
approximately 0.0018 together with an almost negligible amount of thrust compression
strain. This additional strain, when superimposed on the existing strain caused by the
static load, may exceed the compression capacity of the concrete.
It is questionable, however, that designing the lining to conform to the perforated
ground deformation (Equation 4-4) is adequate in this case. Flexibility ratio equal to 1.0
implies that the lining may just have enough stiffness to replace that of the soil being
excavated. Ideally, the lining should distort in accordance with the free-field, non-perforated
ground deformation (Equation 4-3). With this assumption, the maximum
diametric strain according to Equation 4-3 is 0.004, a value only one-third of that
calculated by Equation 4-4.
A computation by finite difference code is performed for comparison. The resulting
maximum diametric strain is about 0.0037, which supports the suggestions made
immediately above.
Summary and Conclusions
In conclusion, the simplified seismic design approach can serve its purpose, provided
that good judgment is used during the design process. The ovaling effects on the lining,
however, may in some cases be overestimated or underestimated, depending on the
relative stiffness between the ground and the lining. The main reason for this drawback is
the uncertainty of the tunnel-ground interaction.
This drawback, however, may be immaterial for most applications in the real world.
For most circular tunnels encountered in practice, the flexibility ratio, F, is likely to be large
enough (F>20) so that the tunnel-ground interaction effect can be ignored (Peck, 1972).
In these cases, the distortions to be experienced by the lining can be reasonably assumed
to be equal to those of the perforated ground.
63

This rule of thumb procedure may present some design problems in the real world
too. These problems arise when a very stiff structure is surrounded by a very soft soil. A
typical example would be to construct a very stiff immersed tube in a soft lake or river bed.
In this case the flexibility ratio is very low, and the tunnel-ground interaction must be
considered to achieve a more efficient design.
In the following section a refined procedure, equally simple, if not simpler, will be
presented. This refined procedure considers the tunnel-ground interaction effect and
provides a more accurate assessment of the seismic effect upon a circular lining.
4.5 Lining-Ground Interaction
Closed Form Solutions
Closed form solutions for estimating ground-structure interaction for circular tunnels
have been proposed by many investigators. These solutions are commonly used for static
design of tunnel lining. They are generally based on the assumptions that:
• The ground is an infinite, elastic, homogeneous, isotropic medium.
• The circular lining is generally an elastic, thin walled tube under plane strain
conditions.
The models used in these previous studies vary in the following two major
assumptions, the effects of which have been addressed by Mohraz et al. (1975) and
Einstein et al. (1979):
• Full-slip or no-slip conditions exist along the interface between the ground and the
lining.
• Loading conditions are to be simulated as external loading (overpressure loading) or
excavation loading.
Most of the recent developments in these models fall into the category of excavation
loading conditions, as they represent a more realistic simulation of actual tunnel
excavation (Duddeck and Erdmann, 1982). To evaluate the effect of seismic loading,
however, the solutions for external loading should be used. Peck, Hendron, and Mohraz
(1972), based on the work by Burns and Richard (1964) and Hoeg (1968), proposed
closed form solutions in terms of thrusts, bending moments and displacements under
external loading conditions.
64

The expressions of these lining responses are functions of flexibility ratio and
compressibility ratio as presented previously in Equations 4-5 and 4-6. The solutions also
depend on the in-situ overburden pressure, gtH, and the at rest coefficient of earth
pressure, Ko. To be adapted to the loading caused by seismic shear waves, it is
necessary to replace the in-situ overburden pressure with free-field shear stress, t, and
assign Ko=–1, to simulate the simple shear condition in the field. The shear stress, t, can
be expressed as a function of shear strain, g. With some mathematical manipulations, the
resulting expressions for maximum thrust, Tmax, bending moment, Mmax, and diametric
strain, DD/D, can be presented in the following forms:
T max =±
1
6
K1
Em
(1 +vm)
Rgmax
Mmax = ±
1
6
K1
Em
(1 +vm)
R2gmax
DD
D
= ±
1
3
K1Fg max
K1 =
12(1 -vm)
2F +5 -6vm
where (Eq. 4-11)
where Em, nm = modulus of elasticity and Poisson’s Ratio of medium
gmax = maximum free-field shear strain
F = flexibility ratio
(Eq. 4-8)
(Eq. 4-9)
(Eq. 4-10)
K1 is defined herein as lining response coefficient. The earthquake loading parameter
is represented by the maximum shear strain, gmax, which may be obtained through a
simplified approach (such as Equation 4-1), or by performing a site-response analysis.
To ease the design process, Figures 10 and 11 show the lining response coefficient,
K1, as a function of flexibility ratio and Poisson’s Ratio of the ground. It should be noted
that the solutions provided here are based on the full-slip interface assumption.
65

Figure 10.
Lining Response Coefficient, K1
(Full-Slip Interface)
66
Response Coefficient, K1

67
Response Coefficient, K1
Figure 11.
Lining Response Coefficient, K1

Comments on Closed Form Solutions
According to previous investigations, during an earthquake slip at interface is a
possibility only for tunnels in soft soils, or when seismic loading intensity is severe. For
most tunnels, the condition at the interface is between full-slip and no-slip. In computing
the forces and deformations in the lining, it is prudent to investigate both cases and the
more critical one should be used in design. The full-slip condition gives more conservative
results in terms of maximum bending moment, Mmax , and lining deflections DD.
This conservatism is desirable to offset the potential underestimation (10 to 15
percent) of lining forces resulting from the use of equivalent static model in lieu of the
dynamic loading condition (Mow and Pao, 1971). Therefore, the full-slip model is adopted
for the present study in evaluating the moment and deflection response of a circular tunnel
lining.
The maximum thrust, Tmax, calculated by Equation 4-8, however, may be significantly
underestimated under the seismic simple shear condition. The full-slip assumption along
the interface is the cause. Therefore, it is recommended that the no-slip interface
assumption be used in assessing the lining thrust response. The resulting expressions,
after modifications based on Hoeg’s work (Schwartz and Einstein, 1980), are:
(Eq. 4-12)
where the lining thrust response coefficient, K2 , is defined as:
È 2
Î
F = flexibility ratio as defined in Eq. 4-6
C = Compressibility ratio as defined in Eq. 4-5
Em, nm = modulus of elasticity and Poisson’s Ratio of medium
tmax = maximum free-field shear stress
gmax = maximum free-field shear strain
K2 =1 +
F[(1-2nm)-(1 -2nm)C]-
1
2
(1-2nm)2 +2
F[(3 -2nm)+(1 -2nm)C]+C
5
2
-8nm +6nm
˘
°
+6 -8nm
T max =±K2tmax R
=±K2
Em
2(1+ nm)
Rg max
68

69
Thrust Response Coefficient, K2
Figure 12.
Lining Response (Thrust) Coefficient, K2
(No-Slip Interface)

Figure 13.
(No-Slip Interface)
70

71
Figure 14.
(No-Slip Interface)

A review of Equation 4-12 and the expression of K2 suggests that lining thrust
response is a function of compressibility ratio, flexibility ratio and Poisson’s Ratio. Figures
12 through 14 graphically describe their interrelationships. As the plots show:
• The seismically induced thrusts increase with decreasing compressibility ratio and
decreasing flexibility ratio when the Poisson’s Ratio of the surrounding ground is less
than 0.5.
• When the Poisson’s Ratio approaches 0.5 (e.g., for saturated undrained clay), the
lining’s thrust response is essentially independent of the compressibility ratio.
Figures 12 through 14, along with data contained in Figures 10 and 11 provide a
quick aid for designers. The theoretical solutions and the influence of interface
assumptions will be further verified for their reasonableness by numerical analysis
presented in the next section.
Another useful and important information, for illustration purpose, is to express the
deformation ratio between the lining and the free-field as a function of flexibility ratio, F.
This relationship can be obtained by dividing Equation 4-10 with Equation 4-3. The
resulting expression is:
(Eq. 4-13)
DDlining
DDfree - field
=
2
3
K1F
The normalized lining deflection is plotted and presented in Figures 15 and 16.
The results indicate that the lining tends to resist and therefore deforms less than the
free-field when the flexibility ratio, F, is less than approximately 1. This situation may occur
only when a stiff lining is built in soft to very soft soils. As the flexibility ratio increases, the
lining deflects more than the free-field and may reach an upper limit as the flexibility ratio
becomes infinitely large. This upper limit deflection is equal to the perforated ground
deformations calculated by Equation 4-4, signaling a perfectly flexible lining situation. The
relationship shown in Figures 15 and 16 supports and supplements the discussions
presented in Examples 1 and 2 of Section 4.3.
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73
Figure 15.
Normalized Lining Deflection

Figure 16.
Normalized Lining Deflection
74
(DDlining)/(DDfree-field)

75
Figure 17.
Finite Difference Mesh
(Pure Shear Condition)

Numerical Analysis
A series of computer analyses using finite difference code (FLAC, 1989) is performed
to verify the proposed procedure in the previous section. The mesh and the lining-ground
system used in these analyses are shown in Figure 17. The assumptions made for these
analyses include the following:
• Plane strain conditions are assumed.
• Seismic shear wave loading is simulated by pure shear conditions with shear stresses
applied at far external boundaries.
• Taking advantage of the anti-symmetric loading conditions, only one quarter of the
entire lining/ground system is analyzed. Rollers are provided at planes of anti-symmetry.
• Lining is modeled by a series of continuous flexural beam elements of linear elasticity.
• Ground (medium) is modeled as linear elastic material.
• No-slip condition along the lining-ground interface is assumed.
A total of 13 analyses are performed. In order to cover a wide range of possible
effects of lining-ground interaction, the parameters for lining and ground are varied.
Following is a list of the range of the variations:
Range of Em = from 325 ksf to 72000 ksf
nm = 0.25 and 0.333
El/(1-nl
2) = 518400 ksf and 662400 ksf
Range of t = from 0.5 feet to 2.0 feet
The resulting flexibility ratios, F, and compressibility ratios, C, are tabulated in Table 2.
To make the level of seismic loading within a reasonable range, the boundary shear
stresses (tmax) are made to result in the maximum free-field shear strains (gmax) in the
range between 0.001 and 0.008.
Results and Recommendations
Maximum Bending Moment, Mmax . The resulting maximum bending moments are first
calculated for each of the 13 cases by using the full-slip closed form solution, Equation 4-
9. These values are then compared to those obtained from the no-slip finite difference
analysis. A plot of comparison in terms of dimensionless bending moment between the
two is shown in Figure 18. As expected, the full-slip interface assumption results in higher
76

77
Table 2.
Cases Analyzed by Finite Difference Modeling

Figure 18.
Influence of Interface Condition on Bending Moment
78

maximum bending moment than the no-slip interface condition. The differences are within
approximately 20 percent under seismic shear loading condition.
It should be realized, however, that these results are based on pseudo-static solutions
that do not consider the potential dynamic amplification and stress concentrations at the
tunnel excavation boundary (Mow and Pao, 1971). Previous studies suggest that a true
dynamic solution would yield results that are 10 to 15 percent greater than an equivalent
static solution, provided that the seismic wavelength is at least about 8 times greater than
the width of the excavation (cavity).
Therefore, it is prudent to adopt the more conservative full-slip assumption for the
calculation of bending moments. With this more conservative assumption, the effects of
stress amplification need not be considered.
Maximum Lining Deflection, D Dlining. Figure 19 presents a plot of the maximum lining
deflections from full-slip closed form solution versus those from no-slip finite difference
analysis (noting that these lining deflections are normalized with respect to the free-field
ground deflections). Similar to the discussion presented above, lining tends to oval
(distort) more under the full-slip interface assumption. The differences, however, are very
small.
The full-slip assumption (Equation 4-10 or Equation 4-13) is recommended for
calculating the lining distortion. The effects of stress amplification need not be considered
when the conservative full-slip assumption is adopted.
It is interesting to note from the plot that almost no difference exists between the two
assumptions for Case No. 12. This can be explained by the fact that a nearly “perfectly
flexible” lining is used and little lining-ground interaction is involved in the Case No.12
analysis.
Maximum Lining Thrust, Tmax. For comparison, the maximum lining thrusts are
calculated using closed form solutions for both assumptions (Equations 4-8 and 4-12).
The results, along with those from the finite difference analysis, are tabulated in Table 3.
The table shows excellent agreement on the thrust response between the numerical finite
difference analysis and the closed form solution for the no-slip condition. It also verified
that the full-slip assumption will lead to significant underestimation of the lining thrust under
seismic shear condition.
Therefore, it is recommended that Equation 4-12 be used for thrust calculation. To
account for the dynamic stress amplification due to the opening, it is further recommended
that thrusts calculated from Equation 4-12 be multiplied by a factor of 1.15 for design
purpose.
79

Figure 19.
Influence of Interface Condition on Lining Deflection
80

81
Table 3.
Influence of Interface Conditions on Thrust

Lining Stiffness, I. The results presented above are based on the assumption that the
lining is a monolithic and continuous circular ring with intact, elastic properties. Many
circular tunnels are constructed with bolted or unbolted segmental lining. Besides, a
concrete lining subjected to bending and thrust often cracks and behaves in a nonlinear
fashion. Therefore, in applying the results presented herewith, the effective (or, equivalent)
stiffness of the lining will have to be estimated first. Some simple and approximate
methods accounting for the effect of joints on lining stiffness can be found in the literature:
• Monsees and Hansmire (1992) suggested the use of an effective lining stiffness that is
one-half of the stiffness for the full lining section.
• Analytical studies by Paul, et al., (1983) suggested that the effective stiffness be from
30 to 95 percent of the intact, full-section lining.
• Muir Wood (1975) and Lyons (1978) examined the effects of joints and showed that
for a lining with n segments, the effective stiffness of the ring was:
(Eq. 4-14)
where Ie < I and n > 4
Ie = Ij +
4
n
Ê
Ë
ˆ¯
2 I
I =lining stiffness of the intact, full-section
Ij = effective stiffness of lining at joint
Ie = effective stiffness of lining
82

5.0 RACKING EFFECT ON RECTANGULAR TUNNELS
83

5.0 RACKING EFFECT ON RECTANGULAR TUNNELS
This chapter first addresses some of the conventional methods used in seismic
racking design of cut-and-cover tunnels and the limitations associated with these
methods. To provide a more rational design approach to overcoming these limitations, an
extensive parametric study was conducted using dynamic finite-element soil-structure
interaction analyses.
The purpose of these complex and time consuming analyses was not to show the
elegance of the mathematical computations. Neither are these complex analyses
recommended for a regular tunnel design job. Rather, they were used to generate sets of
data that can readily be incorporated into conventional design procedures. At the end of
this chapter, a recommended procedure using simplified frame analysis models is
presented for practical design purposes.
5.1 General
Shallow depth transportation tunnels are often of rectangular shape and are often built
using the cut-and-cover method. Usually the tunnel is designed as a rigid frame box
structure. From the seismic design standpoint, these box structures have some
characteristics that are different from those of the mined circular tunnels, besides the
geometrical aspects. The implications of three of these characteristics for seismic design
are discussed below.
First, cut-and-cover tunnels are generally built at shallow depths in soils where seismic
ground deformations and the shaking intensity tend to be greater than at deeper locations,
due to the lower stiffness of the soils and the site amplification effect. As discussed in
Chapter 2, past tunnel performance data suggest that tunnels built with shallow soil
overburden cover tend to be more vulnerable to earthquakes than deep ones.
Second, the dimensions of box type tunnels are in general greater than those of
circular tunnels. The box frame does not transmit the static loads as efficiently as the
circular lining, resulting in much thicker walls and slabs for the box frame. As a result, a
rectangular tunnel structure is usually stiffer than a circular tunnel lining in the transverse
direction and less tolerant to distortion. This characteristic, along with the potential large
seismic ground deformations that are typical for shallow soil deposits, makes the soil-structure
interaction effect particularly important for the seismic design of cut-and-cover
rectangular tunnels, including those built with sunken tube method.
85

Third, typically soil is backfilled above the structure and possibly between the in-situ
medium and the structure. Often, the backfill soil may consist of compacted material
having different properties than the in-situ soil. The properties of the backfill soil as well as
the in-situ medium should be properly accounted for in the design and analysis.
5.2 Racking Effect
During earthquakes a rectangular box structure in soil or in rock will experience
transverse racking deformations (sideways motion) due to the shear distortions of the
ground, in a manner similar to the ovaling of a circular tunnel discussed in Chapter 4. The
racking effect on the structure is similar to that of an unbalanced loading condition.
The external forces the structure is subjected to are in the form of shear stresses and
normal pressures all around the exterior surfaces of the box. The magnitude and
distribution of these external forces are complex and difficult to assess. The end results,
however, are cycles of additional internal forces and stresses with alternating direction in
the structure members. These dynamic forces and stresses are superimposed on the
existing static state of stress in the structure members. For rigid frame box structures, the
most critical mode of potential damage due to the racking effect is the distress at the top
and bottom joints.
Damages to shallow buried cut-and-cover structures, including regular tunnel
sections, were reported during the earthquakes of 1906 San Francisco and 1971 San
Fernando (Owen and Scholl, 1981). The damages included:
• Concrete spalling and longitudinal cracks along the walls
• Failure at the top and bottom wall joints
• Failure of longitudinal construction joints
For structures with no moment resistance — such as the unreinforced brick arch in
one of the cases during the 1906 San Francisco earthquake — total collapse is a
possibility.
The methods used in current design practice to counteract the seismic effects on
rectangular tunnel linings are described in the following two sections (5.3 and 5.4).
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5.3 Dynamic Earth Pressure Methods
Mononobe-Okabe Method
Dynamic earth pressure methods have been suggested for the evaluation of
underground box structures by some engineers. The most popular theory for determining
the increase in lateral earth pressure due to seismic effect is the Mononobe-Okabe theory
described, for example, by Seed and Whitman (1970), recognized by Japanese Society of
Civil Engineers for earthquake resistant design of submerged tunnels (1975), and
recommended in several other documents (Converse Consultants, 1983; EBMUD, 1973).
Using this method, the dynamic earth pressure is assumed to be caused by the inertial
force of the surrounding soils and is calculated by relating the dynamic pressure to a
determined seismic coefficient and the soil properties.
Originally developed for aboveground earth retaining walls, the Mononobe-Okabe
method assumes that the wall structure would move and/or tilt sufficiently so that a yielding
active earth wedge could form behind the wall. For a buried rectangular structural frame,
the ground and the structure would move together, making it unlikely that a yielding active
wedge could form. Therefore, its applicability in the seismic design of underground
structures has been the subject of controversy.
The obvious applicable situation is limited to the typical “boat section” (i.e., U-section)
type of underground construction, where the structure configuration resembles that of
conventional retaining walls. Another situation where the use of the Mononobe-Okabe
method may also be adequate is when the structure is located at a very shallow depth.
Experience from PB’s recent underground transportation projects has indicated that the
Mononobe-Okabe earth pressure, when considered as an unbalanced load, may cause a
rectangular tunnel structure to rack at an amount that is greater than the deformation of the
surrounding ground. This unrealistic result tends to be amplified as the depth of burial
increases. This amplification is primarily due to the inertial force of the thick soil cover,
which acts as a surcharge and, according to the Mononobe-Okabe method, has to be
considered. In spite of this drawback, the method has been shown to serve as a
reasonable safety measure against dynamic earth thrust for tunnels buried at shallow
depths (e.g., in the Los Angeles Metro Project).
Wood Method
Another theoretical form of dynamic earth pressure was derived by Wood (1973). By
assuming infinite rigidity of the wall and the foundation, Wood derived a total dynamic
87

thrust that is approximately 1.5 to 2.0 times the thrust calculated by the Mononobe-Okabe
method. Model experiments by Yong (1985) confirmed these theoretical results. This
method is possibly adequate for a volume structure (e.g., a basement) resting on a very
stiff/hard medium (such as rock) and rigidly braced across (e.g., by transverse shear wall
diaphragms). A possible application of this method in a cut-and-cover tunnel construction
is at the end walls of a subway station, where the end walls act as rigid shear wall
diaphragms and prevent the structure from making sideways movements during
earthquakes. For regular rectangular cross-sections under plane strain condition, the
Wood theory, like the Mononobe-Okabe method, would lead to unrealistic results and is
not recommended for use in typical tunnel sections with significant soil cover thickness.
Implications for Design
It is logical to postulate that the presence of a rectangular frame structure in the
ground will induce dynamic earth pressures acting upon the structure. This earth pressure
loading, however, is in a form of complex distributions of shear stresses as well as normal
pressures along the exterior surfaces of the roof, the walls and the invert. To quantify
these external earth loads accurately requires a rigorous dynamic soil-structure analysis.
Realizing that the overall effect of this complex external earth loading is to cause the
structure to rack, engineers find it more realistic to approach the problem by specifying
the loading in terms of deformations. The structure design goal, therefore, is to ensure that
the structure can adequately absorb the imposed racking deformation (i.e., the
deformation method), rather than using a criterion of resisting a specified dynamic earth
pressure (i.e., the force method). The focus of the remaining sections of this chapter,
therefore, is on the method based on seismic racking deformations.
5.4 Free-Field Racking Deformation Method
Conventionally, a rectangular tunnel structure is designed by assuming that the
amount of racking imposed on the structure is equal to the free-field shear distortions of
the surrounding medium. The racking stiffness of the structure is ignored with this
assumption. In Section 4.2 (Chapter 4), the commonly used approach to estimating the
free-field shear distortions of the medium was discussed. Using the free-field racking
deformation method, Figure 20 shows a typical free-field soil deformation profile and the
resulting differential distortion to be used for the design of a buried rectangular structure.
88

89
Figure 20.
Typical Free-Field Racking Deformation
Imposed on a Buried Rectangular Frame
(Source: St. John and Zahrah, 1987)

San Francisco BART
In his pioneering development of the seismic design criteria for the San Francisco
BART subway stations, Kuesel (1969) presented this approach and developed project-specific
soil distortion profiles for design purpose. The elastic and plastic distortion limits
of the reinforced concrete box structure were studied and compared to the design free-field
soil distortions. For the BART project, Kuesel concluded that:
• The structure would have sufficient capacity to absorb the imposed free-field soil
distortions elastically in most cases, and that no special provisions need be made for
seismic effects.
• When the imposed shear distortions caused plastic rotation of joints, such joints
should be designed with special structural details.
The soil deformation profiles and some of the assumptions used by Kuesel at that time
are applicable only for the SFBART project. The design philosophy and the general
approach proposed are still valid, however, even when viewed more than two decades
later.
Los Angeles Metro
In setting forth the seismic design criteria for the LA Metro project, Monsees and
Merritt (1991), also adopted the free-field deformation method for the racking evaluation of
rectangular frame structures. They specified that joints being strained into plastic hinges
should be allowed under the Maximum Design Earthquake (MDE) provided that no plastic
hinge combinations were formed that could lead to a potential collapse mechanism. The
acceptable and unacceptable hinging conditions specified in the LA Metro project are
described in Figure 21.
Flexibility vs. Stiffness
In contrast to the static design, where the loads are well defined and the analysis is
based on a “force method,” the seismic effect based on the “deformation method” is
highly dependent on the structural details. The seismic forces induced in structural
members decrease as the structure’s flexibility increases. Therefore, from the seismic
design standpoint it is desirable to make the structure flexible rather than to stiffen it. In
90

91
Figure 21.
Structure Stability for Buried Rectangular Frames
(Source: Monsees and Merritt, 1991)

general, flexibility can be achieved by using ductile reinforcement at critical joints. In
contrast, increasing the thickness of the members makes the structure less flexible. The
special structural details suggested by Kuesel and the plastic-hinge design specified by
Monsees and Merritt are in fact based on this philosophy.
Another design concept that can increase the flexibility of the cut-and-cover box
structure is to specify pinned connections at walls/slabs joints. This design detail
becomes attractive when cofferdam retaining structures are used as permanent walls
because pinned connections are less difficult to build than fixed connections in this case.
Applicability of the Free-Field Racking Method
The free-field deformation method serves as a simple and effective design tool when
the seismically induced ground distortion is small, for example when the shaking intensity
is low or the ground is very stiff. Given these conditions, most practical structural
configurations can easily absorb the ground distortion without being distressed. The
method is also a realistic one when the structure, compared to its surrounding medium, is
flexible.
Cases arise, however, when this simple procedure leads to overly conservative
design for box structures. These situations generally occur in soft soils. Seismically
induced free-field ground distortions are generally large in soft soils, particularly when they
are subjected to amplification effects. Ironically, rectangular box structures in soft soils are
generally designed with stiff configurations to resist the static loads, making them less
tolerant to racking distortions. Imposing free-field deformations on a structure in this
situation is likely to result in unnecessary conservatism, as the stiff structure may actually
deform less than the soft ground. An example to demonstrate the effect of structure
stiffness on racking deformation is given below.
Examples
Soil Parameters. In this example a simplified subsurface profile is used in the free-field
deformation analysis and the soil-structure interaction analysis. Figure 22 shows the soil
stratigraphy of this profile. Shear wave velocities are used to represent the stiffness of the
soil layers overlying the bedrock. For parametric study purposes, the analysis is
performed for two cases with the silty clay layer being represented by a shear wave
velocity of:
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93
Figure 22.
Soil-Structure System Analyzed in Example

• 254 ft/sec for case I
• 415 ft/sec for case II
These shear wave velocities are assumed to be compatible with the shear strains the
soil experiences during the design earthquake. Assuming a unit weight of 115 pcf for the
silty clay, the corresponding shear moduli are:
• G = 230 ksf for case I
• G = 615 ksf for case II
Figure 23 shows the shear wave velocity profiles used in the analysis.
Structure Properties. A reinforced one-barrel concrete box structure with the following
properties is assumed:
Structure Elastic* Moment of Thickness Length
Member Modulus(ksi) Inertia(ft4/ft) (ft) (ft)
Side Wall 3640 42.7 8.0 26
Base Slab 3640 51.2 8.5 90
Roof Slab 3640 51.2 8.5 90
* Plane Strain Elastic Modulus
The structure members are modeled as rigid continuous beam elements under a two-dimensional
plane strain condition.
Analytical Model. Earthquake excitation is represented by a vertically propagating shear
wave accelerogram originated from the rigid bedrock. The relative geometric relationship
between the soil and the tunnel structure is described in Figure 22.
To assess the effect of soil-structure interaction the analysis is conducted using the
dynamic finite element program FLUSH (1975). Under horizontal earthquake excitation
the seismic loading condition is anti-symmetrical. Therefore, only one half of the soil-structure
system need be analyzed, by imposing horizontal rollers along the vertical axis of
anti-symmetry (see Figure 21). A more detailed description of the time-history finite
element analysis including the input ground motions and the structural modeling will be
given in Section 5.5.
Results. Figure 24 shows results based on free-field analysis, ignoring the presence of
structure and the opening. The free-field differential deformations between the projected
94

95
Figure 23.
Subsurface Shear Velocity Profiles

locations of roof and invert are approximately 0.26 inch and 0.17 inch for case I and case
II respectively. When both soil and structure are included in the analysis, the calculated
racking distortions (between the roof and the invert) were only about 13 percent and 32
percent of the free-field deformations for case I and case II, respectively (see Figures 25
and 26).
Conclusions. The results of the analysis lead to the following conclusions:
• It may be very conservative to design a rectangular tunnel structure to accommodate
all the shear deformations in the free-field, particularly when the structure is stiff and
the surrounding ground is soft. This finding coincides with results from several
previous studies (Hwang and Lysmer, 1981; and TARTS, 1989).
• As the relative stiffness between the soil and the structure decreases (e.g., from case
II to case I), the actual structure racking deformation would also decrease, when
expressed as a percentage of the free-field deformation. This suggests that the soil-structure
interaction effect on the racking of a rectangular tunnel should be:
-Similar to that on the ovaling of a circular tunnel (Chapter 4)
-A function of the relative stiffness between the ground and the structure
A series of analyses performed to define this relationship and their results are
presented and discussed next.
5.5 Tunnel-Ground Interaction Analysis
Although closed-form solutions accounting for soil-structure interaction, such as those
presented in Chapter 4, are available for deep circular lined tunnels, they are not available
for rectangular tunnels due primarily to the highly variable geometrical characteristics
typically associated with rectangular tunnels. Conditions become even more complex
because most of the rectangular tunnels are built using the cut-and-cover method at
shallow depths, where seismically induced ground distortions and stresses change
significantly with depth.
It is desirable, therefore, that a simple and practical procedure be developed for use
by design engineers that accounts for the soil-structure interaction effect. To that end, a
series of dynamic soil-structure interaction finite element analyses were performed in this
study. The results from these complex analyses were then transformed so that they could
be adapted easily to simple analytical tools used currently in design practice.
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97
Figure 24.
Free-Field Shear Deformation
(from Free-Field Site Response Analysis, SHAKE)

Figure 25.
Structure Deformations vs. Free-Field Deformations, Case I
(from Soil/Structure Interaction Analysis, FLUSH)
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99
Figure 26.
Structure Deformations vs. Free-Field Deformations, Case II

Factors Contributing to the Soil-Structure Interaction Effect
Many factors contribute to the soil-structure interaction effect. In this study, the main
factors that may potentially affect the dynamic racking response of rectangular tunnel
structures are investigated. These factors are:
• Relative Stiffness between Soil and Structure. Based on results derived for circular
tunnels (see Chapter 4), it is anticipated that the relative stiffness between soil and
structure is the dominating factor governing the soil/structure interaction. Therefore, a
series of analyses using ground profiles with varying properties and structures with
varying racking stiffness was conducted for parametric study purpose. A special case
where a tunnel structure is resting directly on stiff foundation materials (e.g., rock) was
also investigated.
• Structure Geometry. Five different types of rectangular structure geometry were
studied, including one-barrel, one-over-one two-barrel, and one-by-one twin-barrel
tunnel structures.
• Input Earthquake Motions. Two distinctly different time-history accelerograms were
used as input earthquake excitations.
• Tunnel Embedment Depth. Most cut-and-cover tunnels are built at shallow depths.
To study the effect of the depth factor, analyses were performed with varying soil
cover thickness.
A total number of 36 dynamic finite element analyses were carried out to account for
the variables discussed above.
Method of Analysis
Computer Program. The dynamic finite element analyses were performed using the
computer code FLUSH (1975), a two-dimensional, plane strain, finite element program in
frequency domain. Besides calculating the internal forces in the structure members,
FLUSH analysis:
• Produces data in the form of maximum relative movements between any two locations
within the soil/structure system being analyzed
• Allows a simultaneous free-field response analysis and compares the relative
movement between any two locations in the soil/structure system and in the free field
100

These features are ideal for this study because design of the tunnel structures is
based on the “deformation method.” A detailed description of this program can be found
in Lysmer, et al. (1975).
Soil-Structure Model. Figure 27 shows the typical soil-structure finite element model
used. The assumptions related to the model were as follows:
• The structure members are modeled by continuous flexural beam elements of linear
elasticity. Structural frames with rigid connections are considered.
• A rigid base underlies the soil (medium) deposit.
• The soil overburden generally consists of a soft layer overlying a stiffer layer. Except
for 7 cases where the top of the stiffer layer is raised to the invert elevation (to study
the effect of stiff foundation), all cases assume the stiffer layer is below the base of the
structure by a vertical distance of at least one time the full height of the structure.
Materials of both layers are linearly elastic.
• No-slip condition along the soil/structure interface is assumed.
• Taking advantage of the anti-symmetric loading condition, only one half the entire
soil/structure system is analyzed. Horizontal rollers are provided at planes of anti-symmetry.
• To minimize the boundary effect on the geometric dissipation of seismic energy, an
energy absorbing boundary is placed at the far side of the mesh (i.e., transmitting
boundary).
Earthquake Accelerograms. The two digitized ground motion accelerograms employed
in the analyses (see Figures 28A and 28B) were generated synthetically from the two sets
of design response spectra presented in Figure 29. The following should be noted:
• The “W. EQ” spectra and the corresponding accelerogram represent the rock outcrop
ground motions that are typical in the western states of the United States. They were
obtained from the San Francisco BART extension project.
• The “N.E. EQ” spectra and the corresponding accelerogram represent rock outcrop
earthquake motions in the northeastern part of the country. They are taken from the
Seismic Design Criteria of Underground Structures for the Boston Central Artery and
Third Harbor Tunnel project (1990).
• Horizontal earthquake accelerograms are input at the rigid base to simulate the
vertically propagating shear waves.
101

As Figures 28A and 28B show, earthquake motions of these two types have very
different frequency characteristics, with the “N.E. EQ” motions displaying significantly
increased high frequency components. The purpose of using two sets of design response
spectra instead of one was to evaluate the effect of ground motion characteristics on
soil/structure interaction.
Note that these design spectra were developed for motions expected at rock outcrop
(ground surface). For motions to be used as rigid base input in the FLUSH analysis, a
suitable modification of ground motion characteristics should be made. This was achieved
in this study by using the one-dimensional site response analysis program SHAKE based on
wave propagation theory. Details of this de-convolution process can be found in Schnabel,
et al.(1972).
Flexibility Ratio for Rectangular Tunnels
Figure 30 shows the five different types of structure configurations that were analyzed.
Note that although the configurations were limited to five types, the racking stiffness of each
structure type was varied further (for parametric studies) by varying the properties of the
structure members (e.g., EI and EA values). Similarly, the stiffness of the surrounding soil,
as represented by shear modulus, was also varied in such a manner that the resulting
relative stiffness between the soil medium and the structure covered a range that was of
interest. This relative stiffness, as represented by the Flexibility Ratio, F, will be defined in
detail in the following paragraphs.
The flexibility ratio for a rectangular tunnel, just as for a circular tunnel, is a measure of
the flexural stiffness of the medium relative to that of the tunnel structure. Under a seismic
simple shear condition, this relative stiffness may be translated into the shear stiffness of the
medium relative to the lateral racking stiffness of the rectangular frame structure.
General Cases. Consider a rectangular soil element in a soil column under simple shear
condition (see Figure 31). Assume the soil element has a width, L, and a height, H, that are
equal to the corresponding dimensions of the rectangular tunnel. When subjected to the
simple shear stress, t, the shear strain (or angular distortion, g) of the soil element is given
by:
(Eq. 5-1)
g =
D
H
where G = shear modulus of soil
=
t
G
D = shear deflection over tunnel height, H
102

103
Figure 27.
Typical Finite Element Model
(from Structure Type 2)

Figure 28A.
West Coast Earthquake Accelerogram
(on Rock)
104
Acceleration (g)

105
Acceleration (g)
Figure 28B.
Northeast Earthquake Accelerogram
(on Rock)

Figure 29.
Design Response Spectra
(West Coast Earthquake vs. Northeast Earthquake)
106

107
Figure 30.
Types of Structure Geometry Used in the Study

Figure 31.
Relative Stiffness Between Soil and a Rectangular Frame
108

The shear (or flexural) stiffness of the soil element is taken as the ratio of the shear
stress to the corresponding angular distortion as expressed by:
(Eq. 5-2)
t
g
=
t
D / H
=G
When the rectangular frame structure is subjected to the same shear stress, t, the
stress can be converted into a concentrated force, P, by multiplying the shear stress by
the width of the structure (P= tL). The resulting expression for the angular distortion of the
structure becomes:
(Eq. 5-3)
g =
D
H
=
P
HS1
=
tL
HS1
where S1 = the force required to cause an unit racking
deflection of the structure
The flexural (or, racking) stiffness of the structure is, therefore, given by:
(Eq. 5-4)
t
g
=
t
D / H
=
S1H
L
The flexibility ratio, F, is obtained by dividing Equation 5-2 by Equation 5-4. The
resulting expression is:
(Eq. 5-5)
F =
GL
S1H
In the expression above, the unit racking stiffness, S1, is simply the reciprocal of lateral
racking deflection, S1=1/D1 caused by a unit concentrated force (i.e., p=1 in Figure 32A).
For a rectangular frame with arbitrary configuration, the flexibility ratio can be determined
by performing a simple frame analysis using conventional frame analysis programs such
as STAAD-III (see Figure 32A). Additional effort required to perform this type of analysis
should be minimal as most of the computer input is readily established for static design.
Special Case 1. For some of the simple one-barrel frames (Figure 32B), it is possible to
derive the flexibility ratio without resorting to computer analysis. The expression of F
109

developed for a one-barrel frame with equal moment of inertia, IL, for roof and invert slabs
and equal moment of inertia, IH, for side walls is given by:
(Eq. 5-6)
F =
G
24
H2L
EIH
+
HL2
EIL
Ê
Ë
ˆ¯
where E = plane strain elastic modulus of frame
G = shear modulus of soil
IL, IH = moments of inertia per unit width for slabs and walls, respectively
Note that the expressions by Equation 5-6 and Equation 5-7 that follow are valid only
for homogeneous, continuous frames with rigid connections. Reinforced framed concrete
structures are examples of this type of construction.
Special Case 2. The flexibility ratio derived for a one-barrel frame with roof slab moment
of inertia, IR, invert slab moment of inertia, II, and side wall moment of inertia, IW , is
expressed as:
(Eq. 5-7)
where
Y =
F =
G
12
HL2
EIR
Y
Ê
Ë
ˆ¯
(1+ a2)(a1 +3a2)2 +(a1+ a2)(3a2 +1)2
(1 +a1 +6a2)2
a1 =
IR
II
Ê
Ë
ˆ¯
and a2 =
IR
IW
Ê
Ë
ˆ¯
E = plane strain elastic modulus of frame
G = shear modulus of soil
IR, II, IW = moments of inertia per unit width
H
L
Implications of Flexibility Ratios. The derivation of the flexibility ratio presented in this
section is consistent with that for the circular tunnels. The theoretical implications are:
• A flexibility ratio of 1.0 implies equal stiffness between the structure and the ground.
Thus, the structure should theoretically distort the same magnitude as estimated for
the ground in the free-field.
110

111
Figure 32.
Determination of Racking Stiffness

• For flexibility ratios less than 1.0, the structure is considered stiff relative to the free-field
and should distort less.
• An infinitely large flexibility ratio represents a perfectly flexible structure. At this state,
the deformed shape of the structure should be identical to that of a perforated ground.
The size and shape of the perforation, of course, should match the structure.
Results of Analysis
Analyses were first performed for 25 cases of soil/structure systems with varying
combinations of soil profile, structure configuration, input ground motion type and flexibility
ratio. Table 4 lists the details of the combinations for all 25 cases. Note that:
• The backfilled overburden thickness (soil cover) used in these analyses was limited to
a range between 15 and 22.5 feet.
• The soil medium surrounding the embedded structure was assumed to be
homogeneous, except for Cases 10, 14 and 15 where a soil profile with linearly
increasing shear modulus with depth was assumed. An average soil shear modulus
taken at the mid-height of the structure was used to represent the soil stiffness and to
calculate the flexibility ratio for these three cases.
For each of the 25 cases, a free-field site response analysis (i.e., with no structure and
no opening in ground) was first performed, followed subsequently by a corresponding
soil/structure interaction analysis. The free-field site response analysis calculated the free-field
shear deformation of the ground, gfree-field, at the depth where the structure was to be
placed, specifically, the differential shear distortion between the projected locations of the
roof and the invert. The corresponding soil/structure interaction analysis then calculates
the actual racking distortion, gs, of the structure.
Racking Coefficient. A racking coefficient, R, defined as the normalized structure racking
distortion with respect to the free-field ground distortion is given as:
gs
gfree-field
R = =
(Eq. 5-8)
Ds
H
Ê
Ë
ˆ¯
Dfree-field
H
Ê
Ë
ˆ¯
=
Ds
D free-field
112

113
Table 4.
Cases Analyzed by Dynamic Finite Element Modeling

where gs = angular distortion of the structure
Ds = lateral racking deformation of the structure
gfree-field = shear distortion/strain of the free-field
D free-field = lateral shear deformation of the free-field
The racking coefficients, R, obtained from the analyses are presented in the last
column of Table 4 for all 25 cases.
Note that the total structural deformation obtained from the finite element analyses
contains a rigid body rotational movement, which causes no distortion to the cross-section
of the structure. Therefore, this portion of the movement is excluded in the calculation of
the structure racking deformation.
Effect of Relative Stiffness. As expected, results of the analyses indicate that the relative
stiffness between the soil medium and the structure has the most significant influence on
the structure response. This is demonstrated in Figure 33, where the structure racking
coefficients, R, are plotted against the flexibility ratios, F.
• When the flexibility ratio approaches zero, representing a perfectly rigid structure, the
structure does not rack regardless of the distortion of the ground in the free-field. The
normalized structure distortion (i.e., R) increases with the increasing flexibility ratio. At
F=1, the structure is considered to have the same stiffness as the ground and
therefore is subjected to a racking distortion that is comparable in magnitude to the
ground distortion in the free field (i.e., Rª1).
• With a flexibility ratio greater than 1.0, the structure becomes flexible relative to the
ground and the racking distortion will be magnified in comparison to the shear
distortion experienced by the ground in the free field. This latter phenomenon is not
caused by the effect of dynamic amplification. Rather, it is primarily attributable to the
fact that the ground surrounding the structure has a cavity in it (i.e., a perforated
ground). A perforated ground, compared to the non-perforated ground in the free
field, has a lower stiffness in resisting shear distortion and thus will distort more than
will the non-perforated ground.
An interesting presentation of these data for rectangular structures is shown in Figures
34 and 35, where the closed-form solutions obtained for the normalized circular lining
deflections (Figure 15 in Chapter 4) are superimposed. Note that the definitions of
flexibility ratio, F, are different.
• For circular tunnels, Equation 4-6 is used.
• For rectangular tunnels, Equation 5-5, 5-6 or 5-7, as appropriate, is used.
114

115
Racking Coefficient, R = Ds/Dfree-field
Figure 33.
Normalized Racking Deflections
(for Cases 1 through 25)

Figure 34.
Normalized Structure Deflections
116
Structure Deformation
Free-Field Deformation

117
Figure 35.

Because the Poisson’s Ratios of the soil used in all the rectangular cases are between
0.4 and 0.48, for comparison, the data for circular tunnels are shown only for Poisson’s
Ratios of 0.4 and 0.5. The figures show excellent consistency in distortion response
between the two distinctly different types of tunnel configurations. Generally speaking, for
a given flexibility ratio the normalized distortion of a rectangular tunnel tends to be less
than that of a circular tunnel by approximately 10 percent.
The results presented above lead to the following conclusions:
• The conventional seismic design practice for rectangular tunnels (see Section 5.4) is
too conservative for cases involving stiff structures in soft soils (specifically, when
F<1.0).
• Designing a rectangular tunnel according to the free-field deformation method leads
to an underestimation of the tunnel response when the flexibility ratio, F, becomes
greater than 1.0. From a structural standpoint, fortunately, this may not be of major
concern in most cases because F>1.0 may imply the medium (soil/rock) is very stiff,
and therefore the free-field deformation can be expected to be small. F>1.0 may also
imply the structure is very flexible so that the structure can, in general, absorb greater
distortions without being distressed.
• From a practical standpoint, the data presented in Figures 34 and 35 can be used for
design purposes. The normalized deflection curves derived for circular tunnels
(Figures 15 and 16) may serve as upper-bound estimates for tunnels with rectangular
shapes. Note that Figures 15 and 16 are based on Equation 4-13 in Chapter 4.
Effect of Structure Geometry. The effect of structure geometry was studied by using five
different types of box structure configurations (Figure 30) in the 25 cases of analyses listed
in Table 4. The results presented in Figure 33, however, clearly demonstrate that:
• The normalized racking deformations are relatively insensitive to the structure
geometry.
• The soil/structure interaction is mainly a function of the relative stiffness between the
soil and the structure, regardless of the variations of structure types.
Effect of Ground Motion Characteristics. The effect of ground motion characteristics on
the normalized racking deformations is negligible. Consider the comparisons of the
following pairs of analyses listed in Table 4:
• Cases 7 and 9 for structure type 2
118

In each pair of analyses, the parameters characterizing the soil/structure system
are identical except for the input ground motions (i.e., the northeastern versus the
western earthquakes). The seismically induced racking distortions of the structures are
much greater under the assumed western design earthquake than the northeastern
design earthquake. However, for the three comparisons made in this study, the
normalized racking response with respect to the free-field, R, is very little affected by
the type of ground motions used in the analysis. For instance, the calculated racking
response coefficients show negligible difference (R=0.445 vs.R=0.448) between cases
22 and 23.
Effect of Embedment Depth. To determine the effect of shallow embedment depth on
the normalized racking response, finite-element analyses were performed using Type 2
structure as an example. Here, the burial depths of the structure were varied. Table 5
presents the cases that were analyzed for this purpose. Note that flexibility ratio, F,
remained the same for all cases. The normalized racking distortions from these analyses
versus the dimensionless depth of burial, h/H, are presented in Figure 36.
Based on the results, it appears that:
• The normalized racking distortion, R, is relatively independent of the depth of burial for
h/H>1.5 (i.e., soil cover thickness equal to structure height). At this burial depth the
structure can be considered to respond as a deeply buried structure.
• For cases where the depth of embedment is less than 1.5, the normalized racking
distortion decreases as the depth of burial decreases, implying that design based on
data presented in Figures 34 and 35 is on the safe side for tunnels with little to no soil
cover.
Effect of Stiffer Foundation. The results of analyses discussed thus far are primarily for
cases involving structures entirely surrounded by relatively homogeneous soil medium,
including soil profiles with linearly increasing stiffness. A frequently encountered situation
for cut-and-cover tunnels is when structures are built directly on the top of geological
strata (e.g., rock) that are much stiffer than the overlying soft soils.
To investigate the effect of stiffer foundation, seven analyses were performed with
varying foundation material properties as well as varying overlying soil properties. Table 6
lists the various parameters used in each of these analyses. The flexibility ratios shown in
Table 6 are based on the overlying soil modulus only. The stiffness of the more competent
foundation material is not taken into account. The calculated racking distortions, as
normalized by the free-field shear deformations, are presented as a function of the
flexibility ratio in Figure 37.
119

Table 5.
Cases Analyzed to Study the Effect of Burial Depth
120

121
Racking Coefficient, R
Figure 36.
Effect of Embedment Depth on Racking Response
Coefficient, R

In comparison with the results shown in Figure 35 it may be concluded that in general,
the presence of a stiffer foundation would result in some, but not significant, increase in the
normalized racking distortion of the structure.
It should be noted, however, that:
• Although the magnitude of this increase is not significant when expressed in a
“normalized” form, the actual impact to the structure may be significantly greater due
to the increased free-field deformations.
• Normally, amplification of shear strains is expected near the zone of interface between
two geological media with sharp contrast in stiffness.
• Care should be taken, therefore, in estimating the free-field shear deformations in a
soft soil layer immediately overlying a stiff foundation (e.g., rock).
5.6 Recommended Procedure: Simplified Frame Analysis
Models
In Section 5.5 the soil-structure interaction effect has been quantified through a series
of dynamic finite-element analyses. Exercises of such complex analyses are not always
necessary. For practical design purposes, a simplified procedure considering the
interaction effect is desirable.
Therefore, a simple, rational and practical way of solving this problem is presented in
this section, based on the data from soil-structure analyses presented in Section 5.5 . By
following this procedure, an engineer equipped with a conventional frame analysis
program (such as STAAD-III) can easily derive the solution for his design task.
Step-by-Step Design Procedure
The simplified frame analysis models shown in Figure 38 are proposed. A step-by-step
description of this procedure is given below:
(a) Characterize the subsurface conditions at the site and determine the soil/rock
properties based on results from field and laboratory investigations.
122

123
Table 6.
Cases Analyzed to Study the Effect of Stiff Foundation

Figure 37.
124

(b) Derive earthquake design parameters. As a minimum, these parameters should
include peak ground accelerations, velocities, displacements, design response spectra,
and possibly the time-history accelerograms for both Maximum Design Earthquake (MDE)
and Operating Design Earthquake (ODE). This work should be carried out by earthquake
engineers with assistance from geotechnical engineers and seismologists.
(c) Conduct a preliminary design of the structure. Size and proportion members of
the structure based on the loading criteria under static loading conditions. Normally,
applicable design codes for buildings and bridges should be used, recognizing that the
structure is surrounded by geological materials rather than a freestanding configuration.
(d) Based on the soil/rock properties from step (a) and the design earthquake
parameters from step (b), estimate the free-field shear strains/deformations of the ground
at the depth that is of interest. Generally:
• For a deep tunnel in a relatively homogeneous medium the simplified Newmark
method, as presented by Equations 4-1 and 4-2, may be used.
• For shallow tunnels, for tunnels in stratified soil sites, or for tunnels sitting on stiff
foundation medium, a simple one-dimensional site response analysis (e.g., SHAKE) is
desirable.
The end results of this step provide the free-field deformation data, Dfree-field, as
depicted in Figure 38.
(e) Determine the relative stiffness (i.e., the flexibility ratio, F) between the free-field
medium and the structure using the properties established for the structure and the
medium in steps (a) and (c) respectively. Equation 5-5, 5-6 or 5-7, as appropriate, may be
used to calculate the flexibility ratio for a rectangular structure.
(f) Determine the racking coefficient, R, based on the flexibility ratio obtained from
step (e), using the data presented in Figures 34 and 35, or Figure 37 as applicable.
(g) Calculate the actual racking deformation of the structure, Ds, using the values of
Dfree-field and R from steps (d) and (f) as follows:
(Eq. 5-9)
Ds = R Dfree- field
(h) Impose the seismically induced racking deformation, Ds, upon the structure in
simple frame analyses as depicted in Figures 38A and 38B.
125

• Pseudo-Concentrated Force Model for Deep Tunnels (Figure 38A). For deeply buried
rectangular structures, the primary cause of the racking of the structure generally is
attributable to the shear force developed at the exterior surface of the roof. Thus, a
simplified pseudo-concentrated force model provides a reasonable means to simulate
the racking effects on a deep rectangular tunnel. Using a conventional frame analysis
program, this may be achieved by applying a horizontal support movement or an
equivalent concentrated force at the roof level.
• Pseudo-Triangular Pressure Distribution Model for Shallow Tunnels (Figure 38B). For
shallow rectangular tunnels, the shear force developed at the soil/roof interface will
decrease as the soil cover (i.e., soil overburden) decreases. The predominant
external force that causes the structure to rack may gradually shift from the shear
force at the soil/roof interface to the normal earth pressures developed along the side
walls. Therefore, for shallow tunnels, the racking deformation, Ds, should be imposed
by applying some form of pressure distribution along the walls instead of a
concentrated force. The triangular pressure distribution is recommended for this
purpose.
Generally, for a given racking deformation, Ds, the triangular pressure distribution
model (Figure 38B) provides a more critical evaluation of the moment capacity of
rectangular structure at its bottom joints (e.g., at the invert-wall connections) than the
concentrated force model (Figure 38A). On the other hand, the concentrated force model
gives a more critical moment response at the roof-wall joints than the triangular pressure
distribution model.
For design, it is prudent to employ both models in the frame analyses. The more
critical results should govern to account for the complex distributions of shear stresses as
well as normal earth pressures along the exterior surfaces of the structures.
(i) Add the racking-induced internal member forces, obtained from step (h), to the
forces due to other loading components by using the loading combination criteria
specified for the project. The loading criteria presented in Chapter 2 (Equations 2-1
through 2-4) are recommended for this purpose.
(j) If the results from step (i) show that the structure has adequate strength capacity
according to the loading combination criteria (for both MDE and ODE), the design is
considered satisfactory and no further provisions under the seismic conditions are
required. Otherwise, proceed to step (k) below.
(k) If the flexural strength of the structure is found to be exceeded from the step (i)
analysis, the structural members’ rotational ductility should be checked. Special design
provisions using practical detailing procedures should be implemented if inelastic
126

127
Figure 38.
Simplified Frame Analysis Models

deformations result. Section 2.4 in Chapter 2 includes a detailed discussion on the
strength and ductility requirements for both MDE and ODE loading combinations.
(l) The structure, including its members and the overall configurations, should be
redesigned if:
• The strength and ductility requirements based on step (k) evaluation could not be met,
and/or
• The resulting inelastic deformations from step (k) evaluation exceed the allowables
(which depend on the performance goals of the structure)
In this case, repeat the procedure from step (e) to step (l), using the properties of the
redesigned structure section until all criteria are met.
Verification of the Simplified Frame Model
The simplified frame models according to Equation 5-8 and Figures 38A and 38B
were performed for Cases 1 through 5 (see Table 4) to verify the models’ validity. The
bending moments induced at the exterior joints of the one-barrel rectangular framed
structure (simplified analyses) were compared to those calculated by the dynamic finite-element
soil/structure interaction analyses (rigorous analyses). The comparisons are
presented, using the concentrated force model, in Figures 39 and 40 for bending
moments at the roof-wall connections and the invert-wall connections, respectively.
Similar comparisons made for the triangular pressure distribution model are shown in
Figures 41 and 42.
As Figures 39 and 40 show, the simplified frame analyses using the concentrated
force model provide a reasonable approximation of the structure response under the
complex effect of the soil/structure interaction. One of the cases, however, indicates an
underestimation of the moment response at the bottom joints (i.e., invert-wall connections)
by about fifteen percent (Figure 40). When the triangular-pressure distribution model is
used, the simplified frame analyses yield satisfactory results in terms of bending moments
at the bottom joints (Figure 42). The triangular-pressure distribution model, however, is not
recommended for evaluation at the roof-wall connections, as it tends to underestimate the
bending moment response at these upper joints (Figure 41).
Through the comparisons made above, and considering the uncertainty and the many
variables involved in the seismological and geological aspects, the proposed simplified
128

129
Figure 39.
Moments at Roof-Wall Connections

Figure 40.
Moments at Invert-Wall Connections
130

131
Figure 41.
Moments at Roof-Wall Connections

Figure 42.
Moments at Invert-Wall Connections
132

frame analysis models shown in Figures 38A and 38B are considered to comprise an
adequate and reasonable design approach to the complex problem.
5.7 Summary of Racking Design Approaches
In summary, four different approaches to analyzing the seismic racking effect on two-dimensional
cut-and-cover tunnel section have been presented in this chapter. Table 7
summarizes the advantages, disadvantages and applicability of these four approaches.
Based on the comparisons made in Table 7, it can be concluded that:
• The simplified frame analysis procedure recommended in Section 5.6 should be used
in most cases.
• The complex soil-structure interaction finite-element analysis is warranted only when
highly variable ground conditions exist at the site and other methods using
conservative assumptions would yield results that are too conservative.
• The dynamic earth pressure methods (e.g., the Mononobe-Okabe method) should be
used to double check the structure’s capacity for tunnels with small soil burial and
with soil-structure characteristics similar to those of aboveground retaining structures
(e.g., a depressed U-section).
133

Table 7.
Seismic Racking Design Approaches
134

6.0 SUMMARY
A rational and consistent methodology for seismic design of lined transportation
tunnels was developed in this study which was mainly focused on the interaction between
the ground and the buried structures during earthquakes. Although transportation tunnels
were emphasized, the methods and results presented here would also be largely
applicable to other underground facilities with similar characteristics, such as water
tunnels, large diameter pipelines, culverts, and tunnels and shafts for nuclear waste
repositories (Richardson, St. John and Schmidt, 1989).
Vulnerability of Tunnel Structures
Tunnel structures have fared more favorably than surface structures in past
earthquakes. Some severe damages — including collapse — have been reported for
tunnel structures, however, during earthquakes. Most of the heavier damages occurred
when:
• The peak ground acceleration was greater than 0.5 g
• The earthquake magnitude was greater than 7.0
• The epicentral distance was within 25 km.
• The tunnel was embedded in weak soil
• The tunnel lining was lacking in moment resisting capacity
• The tunnel was embedded in or across an unstable ground including a ruptured fault
plane
Seismic Design Philosophy
State-of-the-art design criteria are recommended for transportation tunnel design for
the following two levels of seismic events:
• The small probability event, Maximum Design Earthquake (MDE), is aimed at public
life safety.
137

• The more frequently occurring event, Operating Design Earthquake (ODE), is
intended for continued operation of the facility, and thus economy.
Loading combination criteria consistent with current seismic design practice were
established in this study for both the MDE and the ODE.
The proper seismic design of a tunnel structure should consider the structural
requirements in terms of ductility, strength, and flexibility.
Running Line Tunnel Design
Seismic effects of ground shaking on a linear running tunnel can be represented by
two types of deformations/strains: axial and curvature. The following procedures currently
used in quantifying the axial and curvature deformations/strains were reviewed:
• The simplified free-field method (Table 1 equations), which allows simple and quick
evaluations of structure response but suffers the following drawbacks:
- By ignoring the stiffness of the structure, this method is not suitable for cases
involving stiff structures embedded in soft soils.
- The ground strains calculated by simplified free-field equations (see Table 1) are
generally conservative and may be overly so for horizontally propagating waves
travelling in soft soils.
• The tunnel-ground interaction procedure (beam on elastic foundation), which provides
a more realistic evaluation of the tunnel response when used in conjunction with a
properly developed ground displacement spectrum.
Through several design examples presented in Chapter 3, it was demonstrated that
under normal conditions the axial and curvature strains of the ground were not critical to
the design of horizontally or nearly horizontally aligned linear tunnels. Special attention
should be given, however, to cases where high stress concentrations may develop as
follows (Section 3.6):
• When tunnels traverse two distinctly divided geological media with sharp contrast in
stiffness
• When abrupt changes in tunnel cross sectional stiffness are present, such as at the
connections to other structures or at the junctions with other tunnels
• When the ground ruptures across the tunnel alignments (e.g., fault displacements)
138

• When tunnels are embedded in unstable ground (e.g., landslides and liquefiable
sites)
• When tunnels are locally restrained from movements by any means (i.e., “hard spots”)
Ovaling Effect on Circular Tunnels
Ovaling of a circular tunnel lining is caused primarily by seismic waves propagating in
planes perpendicular to the tunnel axis. Usually, the vertically propagating shear waves
produce the most critical ovaling distortion of the lining.
The conventional simplified free-field shear deformation method was first reviewed,
through the use of several design examples in this study, for its applicability and
limitations. Then a more precise, equally simple method of analysis was developed to
assist the design. This method takes into account the soil-lining interaction effects and
provides closed form solutions (Equations 4-9 through 4-13) to the problems.
Numerical finite difference analyses using the computer program FLAC were
performed to validate the proposed method of analysis. A series of design charts (Figures
10 through 16) was developed to facilitate the engineering design work.
Racking Effect on Rectangular Tunnels
The racking effect on a cut-and-cover rectangular tunnel is similar to the ovaling effect
on a mined circular tunnel. The rectangular box structure will experience transverse
sideways deformations when subjected to an incoming shear wave travelling
perpendicularly to the tunnel axis. The most vulnerable part of the rectangular frame
structure, therefore, is at its joints.
Conventional approaches to seismic design of cut-and-cover boxes consist of:
• The dynamic earth pressure method (Section 5.3), originally developed for
aboveground retaining structures. Its applications in the seismic design of
underground structures are limited only to those built with very small backfill cover,
and those with structural characteristics that resemble the characteristics of
aboveground retaining structures (e.g., a depressed U-section).
• The free-field shear deformation method (Section 5.4), which assumes that the racking
deformation of a tunnel conforms to the shear deformation of the soil in the free-field.
139

Use of this method will lead to a conservative design when a stiff structure is
embedded in a soft soil deposit. On the other hand, when the tunnel structure is
flexible relative to the surrounding ground, this method may also underestimate the
seismic racking response of the structure.
A proper design procedure that can avoid the drawbacks discussed above must
consider the soil-structure interaction effect. For this purpose, an in-depth study using
dynamic finite element soil-structure interaction analysis was conducted (Section 5.5). In
this study, many factors that might potentially affect the tunnel response to seismic effects
were examined. The results, however, indicate that the relative stiffness between the soil
and the structure is the sole dominating factor that governs the soil-structure interaction
effect.
Flexibility ratios, F , were defined to represent the relative stiffness between soils and
rectangular structures. Using these flexibility ratios, a well defined relationship was
established between the actual tunnel racking response and the free-field shear
deformation of the ground (Figures 34 and 35). This relationship allows engineers to
perform their design work by using conventional and simple frame analysis programs
without resorting to complex and time consuming finite element soil-structure interaction
analyses. A detailed step-by-step design procedure using these simplified frame analysis
models was given in Section 5.6 of Chapter 5.
140

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147

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